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  1. Dear Fellows I have encountered an issue; draftsmen/architects take projects from clients and make structural drawings without calculations and there are engineers whose stamps are available at printing shops which are stamped for very minimal charges. Structural engineers who are stamping as designer and as vetter are having filthy money without fulfilling their duties. These designers and vetters disregard the profession and engineering community, they do not have any ethics and sense of humanity. PEC do nothing or there is no complain against them. Being engineers and specially as structural engineers this is our responsibility to raise this issue to PEC or any other authorities and more to educate our engineering community that have respect and dignity in the profession, one must not disrespect his profession or let disrespected by such black sheep in the community. If a system/community have the ability to spit the wrong things out it remains alive and working perfectly. when system/community allows everything without check, such community is already dead. Structural engineers do not get projects who are working earnestly due to such people. Main aspect of design safety of humanity is at stack . This is totally against the PEC Code of Ethics. This is against the PEC Code of Conduct. Allah has bestowed us with great knowledge and brain to use for serving humanity and earning rizq e halal. We do use our brains rightly. We should stop this corruption by whatever means we can do.We must spit these non-professionals out of our community. This forum is one of the unique voice of structural engineering community, people comes here to get and share knowledge. I request to all the members that will stand with me on this issue. There is immense need to educate people that structural engineering is a business of life/economy saving, the fees structural engineers charge is nothing in comparison to the cost of lives lost during any incident due to incomplete and false designs or more accurate no designs. Regards
    10 points
  2. UmarMakhzumi

    Pile Design

    *SEFP Consistent Design**Pile Design**Doc No: 10-00-CD-0005**Date: Nov 21, 2017* This article is intended to cover design of piles using Ultimate Limit State (ULS) method. The use of ULS method is fairly new for geotechnical design (last decade). The method is being used in multiple countries now (Canada, Australia etc). The following items shall be discussed: Overview Geotechnical Design of Piles (Compression Loads, Tension Loads and Lateral Loads) Structural Design of Piles (Covering both Concrete and Steel) Connection of Pile with the foundation (Covering both Concrete and Steel) Pile Group Settlement Things to consider 1. Overview Piles provide a suitable load path to transfer super-structure loads to foundation where shallow foundation are not suitable - this can be due to a number of reasons like existing space constraints or suitable soil strata is not present immediately below structure. Other uses can be to meet design requirements like to have reduced settlement etc. This article shall cover the use of straight shaft cast-in-place concrete piles and straight shaft driven steel pipe piles. There are a number of additional piles types like belled concrete piles, precast concrete piles, screw / helical steel piles etc but the discussion to choose a suitable pile type is not in the intended scope of this article. The article is intended to discuss design requirements for straight shaft piles only (both concrete and steel) . The aforementioned topic about pile selection is a very diverse subject and requires a separate discussion on its own. Before I get into the nitty and gritty of pile design, it is important to highlight that as a structural engineer working on pile design, there are a number of parameters that you would require from the geotechnical engineer. Generally, these parameters are provided in the project geotechnical report. Based on those parameters, the geotechnical design of piles is performed first followed by structural design of pile. The next section talks about the geotechnical design of piles. 2. Geotechnical Design of Piles. Geotechnical design of pile means sizing of pile. This includes determining the following two geometric properties of piles: 1) Diameter or radius 2) Length Straight shaft piles embeded in soil derive their capacity from two sources. The first one is the skin friction along the pile length and the second one is the end bearing. In order to complete the geotechnical design of piles or in simple words to "size up the piles", you will need skin friction values for different soil strata through which the pile would penetrate or lie and the bearing capacity of the layer in which pile would terminate. This information is provided by the geotechnical engineer in the project geotechnical report. Generally, they would provide a table showing skin friction values of each soil layer for both tensile and compressive loads along with end bearing values of each layer. In addition to this, for areas susceptible to frost loading, the geotechnical engineer would also provide ad-freeze and frost heave forces. You can't design a pile without knowing what these values are. So this is something that you need from a geotechnical engineer. Once you have received the project geotechnical report with all the required information, you need to start sizing the piles. The easiest way to do it is to create an excel sheet and do preliminary calculations for different standard diameters like 200mm, 324 mm, 406mm, 460mm, 508mm, 610mm, 762mm and 914mm. The geotechnical report shall also provide recommendations if certain top soil layers need to be ignored or not. Example Problem: From your structural analysis, the maximum factored compressive load is 100 kN. and maximum factored tensile load is 50 kN. You need to size a pile (do geotechnical design) to meet that applied load. Sizing piles for geotechnical capacities is simple. Here is the formula for capacity of pile based on skin friction only (ignoring end bearing for simplicity): ULS Geotechnical Pile Axial Capacity: Pi * Pile Diameter * Total Embedment Length of Pile * Skin Friction Value * Resistance Factor Where, Pi= 3.14 Pile Diameter = 2* Radius Total Embedment Length of Pile = Pile Embedment Length - Frost Depth Skin Friction Values = See geotechnical for values Resistance Factor = 0.4 for compression and 0.3 for tension. For, the above problems, lets assume Skin Friction values of 80 kPa for both tension and compression and initial pile size (diameter) of 324 mm, Frost Depth of 3000 mm. For total length of 10m (lets assume a starting length), Total Embedment Length of Pile = 10m - 3m = 7m (Total Length - Frost Depth) ULS Geotechnical Pile Compressive Capacity= 3.14 * (0.324m) * 7m * 80 kPa * 0.4 = 228 kN > 100 kN Okay. ULS Geotechnical Pile Tensile Capacity = 3.14 * (0.324m) * 7m * 80 kPa * 0.3 = 171 kN > 50 kN Okay. The above problem shows you how to calculate the compressive and tensile capacities (also called the axial capacities) of the pile. For lateral capacity, you will need to know the modulus of sub grade information from the geotechnical engineer and use a software like LPILE to see the response against the lateral load. It is important to note that lateral deflection of pile is a service limit state meaning that it should be checked against unfactored loads. Generally, for petrochemical and oil and gas industries, pile service loads are defined as a deflection limit that will depend upon the maximum allowable movement of pile considering an elastic response from soil as well as the maximum movement piping and its attachments can take. Here is a scenario explaining that. For example, your geotechnical engineer recommends a maximum lateral movement of pile to be limited to 6mm so that soil around pile stays elastic. The structure you are designing, has a wind load deflection of 12mm. The pipes and equipment plus their connections shall be designed for 6mm+12mm = 18mm movement of structure. You need to notify piping of this deflection limit and if they are okay, you are good. If they are not, you will have to stiffen up the structure to lower the overall structure deflection and work with piping to see alternate routing for pipe. For pile design, you need to see what diameter pile shall have a capacity at 6mm lateral deflection greater than the applicable horizontal service load. To calculate pile capacity for different pile head movements, you will need to use LPILE or similar software. LPILE shall provide you a graph that would show you that how much a pile would move under applied lateral load or moment. LPILE is very easy to operate. You can look at the program tutorials and work your way through. It will also provide you the analysis results for a pile embeded in soil with soil modelled as springs along the length. This analysis result is important and allows us to see what is the maximum moment and shear developed in pile due to applicable load and based on combined response of soil and pile interaction. If you don't have LPILE, you can ask the geotechnical engineer, to provide you with pile lateral capacity graphs. In this case, you will need to provide the geotechnical engineer with estimated pile sizes, estimated axial and lateral loads, pile head condition (Fixed or Pinned) upfront. The goetech engineer will run the LPILE for you and provide you the graphs that will show the maximum load a pile can take against different lateral displacement values and would also provide the maximum moment due to max lateral load. I have done this on a number of projects and this is standard industry practice. 3. Structural Design of Piles. After completing the geotechnical design of pile, the structural design of pile needs to be performed. In order to do that, you will need to know the maximum moment in pile due to the application of axial and lateral loads. As mentioned above, the easiest way is to use LPILE output as it provides you with deformed shape of the pile along with the maximum moments and shears due to applied loads - the analysis of pile embedded in soil. Using LPILE analysis results, you can use beam-column capacity formulas to design a steel pile or column interaction diagram to design a concrete pile. Beam-Column capacity formulas vary with different codes so therefore I haven't included any example. For steel piles, corrosion allowance should be considered as per the code requirements. Generally its 1.5mm each exposed face so for pipe piles it will be 3mm considering exterior and interior face of the pile. 4. Connection of Pile with the foundation (Covering both Concrete and Steel) The connection of pile and foundation / pile cap is extremely simple for concrete piles. All you need to do is to develop the bars from concrete pile in concrete foundation/ pile cap. For steel piles, similar concept is there, except for you need to weld rebars on top of cap plate. 5. Pile Group Settlement Single pile or pile groups should always be check for settlement. Geotechnical consultant shall be contacted to get guidance on what method should be used. Methods like equivalent raft method or finite element analysis can be carried out to get settlement numbers. 6. Things to Consider For pile group, group effects are generally provided by the geotechnical engineer that can be applied to pile group. The group effects are a function of pile diameter and centre to centre spacing. Pile capacities are reduced if they are spaced closely. For straight shaft piles, rule of thumb is to place them greater or equal centre to center distance of to 3 * diameter of pile. For lateral loads, pile capacities are reduced at 3 * diameter spacing and generally piles need to be spaced at 5 * diameter to have no lateral reduction. Also, straight shaft piles if placed too close might result in pile installation issues. Some piles already installed might heave up if other piles are being installed in close proximity. Impact of pile driving to existing structures should also be considered especially if there is sensitive instrumentation installed in close proximity. Hope this article provides the much needed guidance on pile design. It is written for beginners and a lot of things have been kept simple. Your feedback is more than welcome. Please post any questions should you have. Thanks.
    9 points
  3. I was browsing through my archives are noticed a bunch of articles written by NICEE (National Information Centre of Earthquake Engineering (NICEE) was established in IIT Kanpur with the mandate to empower all stakeholders in the building industry in seismic safety towards ensuring an earthquake resistant built environment. NICEE maintains and disseminates information resources on Earthquake Engineering. It undertakes community outreach activities aimed at mitigation of earthquake disasters. NICEE’s target audience includes professionals, academics and all others with an interest in and concern for seismic safety). The articles are free to publish as long as original content stays unchanged. These articles are good for fresh structural engineers and Civil/ Structural Engineering Students. The best thing about them is that they are only 2 pages and full of images. It literally takes less than 5 min to go through each. EQTip19.pdf EQTip20.pdf How architectural features effect buildings.pdf How buildings twist during earthquakes.pdf How do Beam-Column Joints in RC Buildings Resist Earthquakes.pdf How do Brick Masonry behave during Earthquake.pdf How do Columns in RC Buildings Resist Earthquakes.pdf How do Earthquake Affect Reinforced Concrete Buildings.pdf How Flexibility of Buildings affect their earthquake response.pdf How the ground shakes.pdf How to make building ductile for Good Seismic Performace.pdf How to make Stone Masonry Buildings Earthquake Resistant.pdf How to Reduce Earthquake Effects on Buildings.pdf What are magnitudes and intensity.pdf What are seismic effects on structures.pdf What causes earthquake.pdf What is seismic design philosophy of Buildings.pdf Why are Buildings with Shear Walls Preferred in Seismic Regions.pdf Why are horizontal bands necessary in masonry buildings.pdf Why are Open Ground Storey Buildings Vulnerable in Earthquakes.pdf Why are Short Columns more Damaged During Earthquake.pdf Why is vertical reinforcement required in masonry buildings.pdf Why should Masonry Buildings have simple Structural Configuration.pdf
    9 points
  4. Young engineers often ask about whether to select membrane or thin shell or thick shell for modelling area elements such as RCC slab. I am going to talk about the difference in the results when choosing thin-shell or membrane. This is what makers of ETABS says about the topic. I will quote the following to clear out the structural difference between the two: "Load which is applied to membrane objects transfers directly to supporting structural objects, whereas meshed shell objects have bending stiffness and therefore resist a portion of the load through flexural deformation. As a result, less load will be available to transfer to beams located under a shell, while 100% of the load will transfer through a membrane". Lets see how much can the results differ. Figure below shows the partial framing plan of a level in the building I will discuss how the modelling of S11a as thin-shell or membrane will effect the design shear and moment on the beam B54a, see figure above. The figure below shows the design forces obtained against the gravity loads for the membrane case: The figure below shows the design forces obtained against the gravity loads for the thin-shell case: Based on the results presented above, B54a's design shear increased by 155%, and its design moment by 147% as compared to S11a's behavior as membrane. Now the question, that a Structural engineer needs to ask is: Which result is more appropriate or which result captures the actual behavior more realistically? And that is where you engineering judgement comes into play.
    6 points
  5. Culverts - first determine bending moment (tensile surfaces) and reinforce either one of two ways: See hand sketch below - once you know moment and effective depth of tension reinforcement you can design the joint. 1. U-bars coming in from slab and wall and connecting in corner. 2. or, Top/outer bar bent around the joint. 3. Option 3 is to use L-bars at joint, - see sketch below for all three ways you can detail a culvert corner - your engineering judgement is determine which detail best suits your design! Spacing of reinforcement tend to be closer to limit serviceability stress and hence crack width in concrete to <2mm - for reinforcement durability, and to stop water corroding the reinforcement, and to have a watertight structure; Movement joints are tricky in culverts. Make sure the concrete (or cement) selected is ok for, (i) against corrosive soils and (ii) chemicals in (waste) water.
    6 points
  6. I accidentally came across these useful case studies, which, I would like to share. You can use them if you are working on a commercial or residential building retrofitting project. These case studies provide insight about seismic retrofitting and also on analytical methods, that are used for building assessment. I would also like to give due credit to people who are involved in these studies. All these studies were performed under a US-Pakistani Joint Cooperation Project. The details for the project are. 6-Storey Mixed Use Building in Karachi.pdf 10-Storey Office Building in Karachi.pdf RS-4 Storey Academic Building in Karachi.pdf Five Storey Residential Apartment.docx
    6 points
  7. 6 points
  8. The quick method to check weather the design of slender column is fine or need to increase the column size are as follows; 1- In order to get the moments along column height accurately, divide the column into 2 or 3 segments. 2- Apply modifier. 3- Activate the P-Delta option from define menu. 4- Run the analysis and design with P-Delta and check the Non sway magnification factor i.e. delta(ns) in design summary of the column (see the pic attached). This delta(ns) must be less than 1.4. If this exceeds the limit of 1.4, increase the column size otherwise the column design is OK. Correction by seniors in my above steps will be highly appreciated.
    6 points
  9. IS PAKISTAN PREPARED ENOUGH TO HANDLE THE NEXT "BIG ONE"?? https://www.express.pk/story/968021/ Errata: The magnitude of 2005 Kashmir earthquake is mistakenly typed as 8.6. Actually it was a M7.6 event.
    6 points
  10. Here are my two cents:- 1. General guidance regarding placement of construction joints in RC work has been provided in Section 6.4 of ACI 318-08 and its commentary. Some clarity is given in section 6.4.3, where it is stated that "Construction joints shall be so made and located as not to impair the strength of the structure. Provision shall be made for transfer of shear and other forces through construction joints." For transfer of shear etc through construction joints, reference is made to the ACI Section 11.6.9 that deals with the calculation of shear-friction, at the interface between two concretes cast at different times (beside other situations described in section 11.6.1 of the code). Moreover, Section 6.4.4 suggests that "Construction joints in floors shall be located within the middle third of spans of slabs, beams, and girders. 2. Regarding construction joints in columns, however, Section 6.4 does not provide guidance clearer than that in Section 6.4.6 stating that the "Beams, girders, or slabs supported by columns or walls shall not be cast or erected until concrete in the vertical support members is no longer plastic." And, the commentary section R6.4.6 explains that "Delay in placing concrete in members supported by columns and walls is necessary to prevent cracking at the interface of the slab and supporting member caused by bleeding and settlement of plastic concrete in the supporting member." 3. The support member (referred in previous paragraph) will generally be a column or a wall. And, in a simplified form, Section 6.4.4 & its commentary are advising us NOT to cast beams & slab monolithically with the wall or column, BUT only after the supporting column (or wall) concrete has hardened, in order to avoid plastic cracking at the beam-column (or beam-wall) joint. 4. In our normal field practice (within Pakistan as well as abroad), beams & slabs are cast at least one day after casting of columns or supporting walls. This gap of one day (between casting of column & beam concretes) ensures that the column (or wall) concrete poured one day earlier has hardened (is no longer plastic), thereby avoiding any possibility of plastic cracking (discussed in paragraph 2 above). 5. Now coming to your queries; In general terms, it is preferable to cast the column in one pour.. However, in compelling circumstances it may be done in more than one pour too, subject to certain conditions. Already described in initial paragraphs. This is the normal & IMHO desirable practice, according to ACI code Section 6.4.6. IMO, leaving 9" or 12" column depth below the beam soffit is excessive & undesirable. It should not be more than 1" or 2" in any case. IMO, this practice is based on the requirements of ACI 318-08 (also ACI 318-11) Section 6.4.6. The same requirement is available in ACI 318-14 Section 26.5.7.2 (a) as well. HTH Regards.
    6 points
  11. *Comments/Observations regarding modelling in ETABS* *Doc No: 10-00-CD-0006* *Date: May 06, 2017* Some of the observations made during extraction of results from ETABS (v 9.7.4), for design of reinforced concrete members, are being share in this article., 1) Minimum Eccentricity ETABS always considers the minimum eccentricity for selecting the design moment of columns irrespective of the probable behavior of the column, whether short or long column. See section 10.10.6.5 and its commentary of ACI 318-08 which deals with minimum eccentricity of long columns. You should always check the design moments that ETABS uses for columns if you want to bring down the cost of construction. 2) Unbraced/ Braced Preference If your model has lateral loads, ETABS will give you design moments in column irrespective of its status as braced or un-braced as per ACI 318 criteria. You should investigate if the storey under consideration is braced, or un-braced (10.10.5.2), and decide appropriate design moments of columns. 3) Time Period ETABS has a tendency to select a time period of the building that is considerably less than the value obtained by the approximate method, Method A, of the section 1630.2.2 of UBC 97. To quote the FEMA 451 document: ''Because this formula is based on lower bound regression analysis of measured building response in California, it will generally result in periods that are lower (hence, more conservative for use in predicting base shear) than those computed from a more rigorous mathematical model". So, there is no need to use the value of time period that is lot less than Ta. One should always check the time period used by the software; ETABS can overestimate the seismic force by more than 2 times. Method A gives lower T and higher V, so FEMA 451 has advised not to use the value of time period less than this value even if rigorous analysis gives a lower value. I have seen the results where Etabs have use the value of time period less than Ta; in-fact as low as 0.5Ta, which can increase the base shear two times. (For a complete discussion on time period, please see the following this thread that complements this section). 4) Stiffness Modifiers First thing is related to modelling the bending stiffness of flexural members, for strength level loads, that is representative of their condition near failure. The ACI code specifies the modifier of 0.35 on gross moment of inertia to represent its condition at yielding. Some people say that the factor should be multiplied by 2 to represent the stiffness of T-beam. This approach would be justified if you are not taking into the account the out of plan bending stiffness of slab. But, ETABS does include the out of plane bending stiffness if you have modelled the slab by using shell elements. So, a factor of 0.7 would overestimate the stiffness of your structure in this case, and will lead to under-design. If one has used the modifier of 0.35 in ETABS for beams in beam-slab floor system, then what value should be adopted for slab? It should not be 0.25, as this value has been specified for flat plates and flat sab floor system. If one is using some value of modifier for out of plane bending stiffness on shells, then the share of the bending moment in beams will be reduced accordingly. This approach is correct if one will be providing the reinforcement in column strips of slab. But, if you are providing reinforcement in slab in the direction perpendicular to supports only, i.e. beams, as is the general practice in Pakistan, then you are under-estimating the flexural demand in beams. Now, there is also a question of factors to be used while deciding the amount of reinforcement required in beams, columns and shear walls. If you are using factors 0.35 for beams and shear walls, and 0.7 for columns, then you are finding out the demand in members at the point of yielding, and this conforms to the code. But, this also means that the structure might experience unacceptable cracks widths. So, if you are using 0.35 for calculating the demand at strength-level forces, then you should also perform crack-control-check at service-level loads by using the factor of 1. If you are calculating the strength-level demand with a modifier of 1 for all structural members, after you have decided the location and the number of shear walls with modifier of 0.35, then you are overestimating seismic forces, as you are underestimating the time-period. But, the structural performance will improve. This article is based on my two separate posts regarding the subject matter. You can view the discussion on the items raised above by viewing the following links: 1) http://www.sepakistan.com/topic/2008-issues-in-etabs-results/ 2) http://www.sepakistan.com/topic/2290-modelling-issuesconsideration-in-etabs/ Thanks.
    6 points
  12. Salam, In reviewing one etabs model, I found that my retaining wall was assigned membrane, when removed and replaced with shell this negative time period problem resolved. Thank you
    5 points
  13. *SEFP Consistent Design* *UBC Seismic Drift Limits* *Doc No: 10-00-CD-0003* *Date: June 04, 2013* The goal of this tutorial is to demonstrate how to evaluate building drifts and story drifts using UBC 97 guidelines. The philosophy behind Story Drift Limits is “Deflection Control”; In UBC 97, deflection control is specified in terms of the story drift as a limit on the lateral displacement of one level relative to the level below. The story drift is determined from the maximum inelastic response, ΔM. Let’s start by defining the design-level response displacements. The elastic deflections due to strength-level design seismic forces are called design-level response displacements. These are denoted by ΔS, where the subscript ‘s’ stands for strength design. Design level response displacements are what you get out of your software, when you run analysis. Please note that structural analysis softwares may provide these values in different formats; say a percentage of height or a direct output. Well, to calculate your story drifts, first you need to find maximum inelastic response displacements from your design-level response displacements. The maximum inelastic response displacement is defined as: ΔM = 0.7RΔS Where, R is the structural system coefficient, the subscript ‘m’ in ΔM signifies that we are calculating a maximum value for the deflection due to seismic response that includes inelastic behavior. Seismic drift values are much larger than wind values. UBC uses maximum inelastic response displacements rather than the design level displacements to verify the performance of the building. Seismic drift limits are 2% & 2.5% of the story height for long and short -period buildings. For a floor to floor height of 12 feet the max., allowable inelastic drift value would be 2% of 12 feet= 0.02*12*12 in=2.88 in. For wind for a 12 story height, drift would be L/400=12*12/400 =0.36 inches, A comparison of both wind and seismic drift limits shows that earthquake inelastic displacements are quiet large compared to wind displacements. That is why proper detailing is emphasized in seismic design. When calculating ΔS for seismic, make sure: You have included accidental torsion in your analysis. Use strength design load combinations: 1.2D + 1.0E + 0.5L & 0.9D + 1.0E. You are using cracked section properties for reinforced concrete buildings. Typical values are Icr walls= 0.5EcIg, Beams = 0.5EcI g & for Columns 0.5 - 0.7 EcIg. Use a reliability/ redundancy factor= 1 to calculate seismic forces. Whenever the dynamic analysis procedure of §1631 is used, story drift should be determined as the modal combination of the story drift for each mode. Determination of story drift from the difference of the combined mode displacements may produce erroneous results because maximum displacement at a given level may not occur simultaneously with those of the level above or below. Differences in the combined mode displacements can be less than the combined mode story drift. Example: A four-story special moment-resisting frame (SMRF) building has the following design level response displacements.(See attached Image) R= 7.0, I= 1 Time period= 0.6 sec (See the attached image for Story Information) Calculate: Maximum Inelastic response displacements. Story drift in story 3 due to ΔM. Check story 3 for story drift limit. Maximum Inelastic response displacements ΔM = 0.7RΔS ΔM = (0.7) (7) ΔS = (4.9) ΔS (See the attached image for Maximum Inelastic response displacements) Story drift in story 3 due to ΔM Story 3 is located between Levels 2 and 3. Thus ΔM drift = 5.39 - 3.43 = 1.96 in. Check story 3 for story drift limit. For structures with a fundamental period less than 0.7 seconds, §1630.10.2 requires that the ΔM story drift not exceed 0.025 times the story height. For story 3: Story drift using ΔM = 1.96 in. Story drift limit = 0.025 *(12*12) in = 3.6 in. > 1.96 in. Therefore, Okay.
    5 points
  14. Following are the documents that you can consider for performing the PBD. You can find guidelines for hinge lengths in them as well. ASCE/SEI 41-17 Seismic Evaluation and Retrofit of Existing Buildings An alternative for seismic analysis and design of tall buildings located in Los Angeles by LATBSDC TBI guidelines for performance based seismic design of tall buildings Performance Based Design State of the Practice for Tall Buildings by EERI Nonlinear Structural Analysis For Seismic Design (NIST GCR 10-917-5) PEER/ATC 72-1 report
    5 points
  15. Maximum reinforcement in beams (flexure only members) are controlled by maximum net tensile strain. The above-mentioned limit corresponds to 8% max reo (4% on each side), similar to column. Beams usually have reo less than 2%. If you are trying to put more reo in a beam using ETABS, change that section property from beam to column. I always love the idea of putting beams as columns in ETABS. A frame section in ETABS defined as a 'column' will be designed for flexure in both direction + axial loads, a true interaction diagram. If defined as beam, ETABS wil ignore any axial load (if present) in beam design.
    5 points
  16. Badar (BAZ)

    Beam Column Joint

    *SEFP Consistent Design**Pile Design**Doc No: 10-00-CD-0007**Date: April 16, 2018* 1.1. FUNCTION OF JOINT Beam-column joint must transfer the forces, such as moment, shear and torsion, transferred by the beam to the column so that the structure can maintain its integrity to carry loads for which it is designed. Another function of the beam-column joint is to help the structure to dissipate seismic forces so that it can behave in a ductile manner. 1.2.WHY DO WE CARE During an extreme seismic event, the code-based structure is expected to maintain its load-carrying capacity for gravity loads even after the structure deforms into inelastic range so that it does not pose any life safety hazard. Hence, the joint can go through significant degradation of strength and stiffness, and if it fails in shear, or anchorage, the life-safety objective of code cannot be achieved. 1.3.CONSEQUENCES OF FAILURE 1.4.THINGS TO CONSIDER FOR BEAM COLUMN JOINT Longitudinal bars of beams, or slab, must be able to develop their yield stress, so that the beam/slab can transfer moment to joint. It means that longitudinal bars must have adequate development length for hooked bars. This implies that the size of the column must be such that bars can develop their tensile forces. If bars can transfer moment, they can also transfer shear as far as monolithic construction is concerned. The shear strength of the joint must enable the transfer of moment and shear through it. The joint should be Constructible: Congestion of reinforcement is the main concern. 1.5.DESIGN SHEAR FOR BEAM COLUMN JOINT The design shear for beam-column joint depends upon the relative strength of beam and column at the joint. For the joints part of the special moment resisting frame, the shear force will be the one that corresponds to the development of hinge in the beam because the frame is required to satisfy strong column-weak beam criteria. If it is a knee joint, then joint area must resist the shear equal to the development of tensile force in the beam. The tensile force will be equal to the product of the area of tension steel, yield strength and the factor that represents the overstrength of steel rebar. If it is not a knee-beam-column joint then, the design shear of the joint will be algebraic sum of tensile force in the beam and the column shear. The column shear is the one that is required to keep the joint in equilibrium, i.e the shear corresponding to the development of the probable moment capacity of beams at the joint. For the joints not part of the special moment resisting frames, one needs to investigate whether the beam or column will yield first. For knee joint, if the column is weaker then the beam, the tensile force cannot exceed the moment corresponding to the development of hinge in column 1.6.THE JOINT: Definition and classification Portion of column within deepest beam that frames in to the column (ACI 352-02). ACI 352-02 categorizes joints based on the displacement-demand imposed by connected members. · TYPE 1 (Section 2.1.1 ACI 352-02) These joints possess limited ductility, and hence the connected members are designed for limited ductility. They are used in situations where ductility of structure is not a concern. · TYPE 2 (Section 2.1.2 ACI 352-02) These joints connect members which designed to have sustained strength under large deformations. Joints are also classified based on their location in framing system 1.7.THE JOINT: Design forces The joint is designed for the shear that results from attainment of the flexural strengths of members connected at the joint for type 2 joints. For type 1 joints, same principle is employed, unless the both members are overdesigned and the engineer does not expect both members, i.e. beam and column, to yield under design forces. 1.7.1. FLEXURAL STREGNTHS: TYPE 2 No strength reduction factor is used for computation of flexural strength. Steel stress is multiplied by factor of 1.25 for computation of flexural strength (3.3.4 ACI 352-02). For type 2 joints, the flexural strength of beams needs to be calculated only, as we do not expect the hinge-formation in columns; we will proportion the beam-column assembly of this joint as per strong-column-weak-beam approach. The slab reinforcement within the flange of beam must also be considered for computation of flexural strength of beam if the slab is integrally cast with beam and if the longitudinal reinforcement of slab is anchored (3.3.2 ACI 352-02). For interior connections, and for exterior and corner connections with transverse beams, the portion of slab to be considered as flange should be as per guidelines of section 6.3.2 of ACI 318-14. The effective flange width should not be taken less than 2 times the width of beam. For exterior and corner connections, without transverse beams, the effective flange width should be as per figures below (section 3.3.2 of ACI 352-02). The effective flange width for this case need not be taken more than 1/12th of the span of the beam. 1.7.2. FLEXURAL STREGNTHS: TYPE 1 For type 1 connection, similar procedure as discussed above should be used, if beams are expected to yield before columns. The stress multiplier factor for this type of connection can be taken as 1. The beam reinforcement, if any, as per section 24.3.4 of 318-14, with-in the effective flange width, must be included in determination of flexural strength in addition to the web reinforcement. If columns are expected to yield before beams, the nominal flexural capacity at beam-column joint should be calculated with due consideration given to the axial load on column. The beam moment in that case would be the one required to maintain equilibrium of the connection. If neither the beam, nor column, is expected to yield at factored loads, then the design shear of joint would be based on factored forces, moments and shear, at beam-column interface.
    5 points
  17. ASD is an older method that is still practised by a lot of mechanical engineers to size pressure vessels and tanks. However, for structural engineering, everywhere, LRFD is used. Here is a detailed answer I read once somewhere on the internet and saved it as it was very interesting.
    5 points
  18. Hello, I'd like to contribute to this forum three excel sheets of my design, Nothing too fancy but that's what I'm capable of The first is a singly reinforced concrete beam - one layer. (Don't use the shear design it's not complete) with ACI 318-14 The second is a hollow core pretension-ed slab analysis using PCI to check the stresses limits, design strength and Mcrack The third excel is an analysis method for seismic design of reinforced concrete water containing structures. ( I've done this very recently if you spot any error in it please inform me ) with ACI 350.3-06 and IBC/ ASCE 7-10 Sheets.rar
    5 points
  19. W/Salaam, Free vibrations are vibrations of the structure (or any other thing) that result once the structure is excited and the vibrations are let to die down without additional interference. For example if you have a pendulum at rest, you pick the bob in your hand, move it on one side and let it go. Once you let go the bob, the pendulum vibrates around the initial rest point. The period of vibration is dependent on the length of pendulum (technically the parameter that defines its stiffness) and is independent of the mass of the bob. If you measure the time the pendulum would take even as it slows down to complete a cycle, it is exactly the same. Amazing isn't it. This is an example of free vibrations. Now, forced vibrations are different. In forced vibration, you have a situation where the load producing or causing vibration of structure (or any other thing) is "continuously being applied" at a frequency. An example would be a reciporcating or centrifugal machine sitting on say concrete foundation. Operation of machine would put the foundation in vibration (machine load itself has a frequency which may or may not match with foundation natural frequency) and would force the foundation response under its frequency. The foundation will as a result vibrate based on its stiffness (like pendulum). Another example considering the case of pendulum (so that you can relate) for forced vibrations would be if you put a big fan infront of the pendulum and it vibrates under the wind of fan. Summary: In free vibrations, you disturb a structure and let it vibrate without disturbing it further. In forced vibrations, the structure is continuously being disturbed by a force applied at regular interval. That is why they are called forced vibrations as we are forcing structural response under external cyclic load. Enjoy the attached forced vibrations video as well that shows breaking of a wine glass under sound. Sound waves are what are forcing the glass to vibrate and the frequency of the sound waves has been set so that its close to natural frequency of the glass causing it to resonate and break.
    5 points
  20. Dear Saiful Islam, Immediate deflection vs long-term deflection is one issue. Linear analysis vs non-linear analysis is another. In ETABS, generally we are almost always performing linear analysis. That means deflection is directly related to the load applied without long-term non-linear effects. So let's ignore non-linear analysis for a moment. In linear analysis, now the question is how do we get immediate and long-term deflection. Immediate deflection could be non-cracked deflection or cracked deflection. Practically speaking, long-term deflection will always be on cracked sections. Although anything "cracked" is essential a non-linear problem, but here let's stick to out point of doing linear analysis only. Now, generally we are not interested in immediate non-cracked deflection. This leaves us with cracked - short-term & long-term deflections in LINEAR analysis. As you are doing LINEAR analysis in ETABS, you cannot get cracked deflections directly. For this, you may apply stiffness reduction modifier option available in ETABS. As you are not applying any stiffness reduction modifier, that means, its NON-CRACKED. This is the answer to your question. However, exploring further, if you want to obtain long-term deflection, you can multiply the immediate deflection by a long-term factor given in ACI chapter 9. ----------------------------- Now comes the non-linear analysis: You can use SAFE for non-linear analysis and get long-term cracked deflections (non-linear) without using modifiers. The cracked non-linear analysis in SAFE will give you long-term cracked deflection which will depend on the extent of cracking which will depend on the reinforcement which will depend on the moment applied. This moment again will depend on the stiffness of cracked section including the reinforcement. So you see, its a non-linear iteration problem. To read further about, linear vs non-linear and elastic vs inelastic analysis, please refer to the book 'Modelling for structural analysis' by Powell.
    5 points
  21. You can do like this; What you have to do is to note down the z coordinate of each shell element along the height (in Excel e.g.) and calculate the force at top and bottom node of each element, then apply the average pressure in local axis 3 (plus or minus). Tip: Always model retaining walls/swimming pool etc so that all the walls have local 3 axis either inside the pool/container or outside. So you can select all the walls once and apply the pressure in one go. And for that turn on 3d view in XZ or YZ in ETABS and select top most mesh, apply pressure and so on. Its not that difficult. To calculate average pressure you can either make your own excel sheet or use the following I once made. Water Pressure on Walls in ETABS.xlsx
    5 points
  22. Dear Engineer, KPK Seismic Field Practicing Manual is attached. Field Practice Manual on basics of good construction practices.pdf Thanks.
    5 points
  23. UmarMakhzumi

    Raft Modifier

    I have worked with some engineers that like to assign high stiffness modifiers to rafts to get conservative flexural and shear design of foundation. Assigning modifiers would increase amount of rebar in your raft. I personally think that this is good practice as nothing is perfectly rigid and cracking in inevitable, which would result in loss of inertia and high flexural stresses. The general procedure is to create two models. One with no stiffness modifiers and one with modifiers for foundation/ raft. You should use the first model to calculate piles reactions and the second to do flexural and shear design. Thanks.
    5 points
  24. In my understanding these are NOT 2 different methods; This is just a differentiation; There are two torsions; 1. Compatibility torsion (where redistribution of moments take place) like slab on beams 2. Equilibrium torsion (where there is no path available for redistribution of moments, like a cantilever slab resting on a beam) These are not two different methods of analysis in ACI or ETABS. This is just to distinguish the cases. That is why it does not matter in ETABS because in ETABS loads will follow the paths that is available. So does not matter if it is case 1 or 2, apply J modifiers but watch for slab moments. Also make sure your detailing handles all these issues. For example if the beam is torsionally too stiff as compared to slab, it will take more moment as compared to slab, and if you are applying less J modifier to beam then make sure the detailing also follows the same approach. (try to increase bottom reinforcement of slab).
    5 points
  25. I have attached tutorials that will show you how to setup ETABS model for dynamic analysis. Go through all tutorials. There are lucid and self explanatory. OneDrive-2014-03-09.zip
    5 points
  26. FEMA P-2012/2018 Assessing Seismic Performance of Buildings with Configuration Irregularities is now available. Thanks. FEMA_P-2012_508.pdf Source: https://www.fema.gov/media-library-data/1551300980344-8e2d825576db50c85ea48448ede5bd90/FEMA_P-2012_508.pdf
    4 points
  27. Mapped Spectral Accelerations Ss and S1 for 43 cities of Pakistan as per The Building codes of Pakistan 2021. Coordinates of cities have been marked on maps and respective accelerations values are tabulated. This task has been performed during the MSC structural Engineering course "Seismic Design of Structures" at UET Lahore, and was assigned by course instructor Prof.Dr.M.Burhan Sharif,CED UET Lahore. Mapped Spectral Acc.Engr.Sameer.UET.pdf
    4 points
  28. Divide option is not meant for meshing in ETABS. It is to "divide" an already drawn panel after one has changed his/her mind regarding the assignments of that element. Having said that, you can use it for meshing as well. But, it will become very cumbersome and lengthy when you want to change some assignments in divided slab elements, as you will then have to select those small panels by zooming on screen. Use meshing option to mesh the area elements. Meshing can be viewed by activating the "Shell Analysis Mesh" option from "Set View Options" tab.
    4 points
  29. Consult PCA notes on ACI 318; chapter 10 of PAC Notes on ACI 318-08. It is better to check them in software by considering the creep and shrinkage as well as excluding the deflections occurring before the installation of partitions.
    4 points
  30. From the seismic point of view, the former is recommended by the code (ACI 318-19). See the attached:
    4 points
  31. Waqar Saleem

    Coding!!!

    Salam, Learning coding is interesting. Instant guidance and help is available, i encountered a site for that freecodecamp, have a look who are interested, wonderful !!!
    4 points
  32. Good points made by BAZ & Umar. Just to add, there is more than one way to skin an animal, as they say – here is another approach to take. The third option would be not to put in "releases" and to "detail" the beam appropriately for torsion - this means the longitudinal bars in the beam may increase in size, and the links (stirrups) are detailed as torsion links. Transferring “releases” from computer software to the steel fixer on site can be tricky. Concrete behaves as you detail it! It is the combination of vertical shear and torsion that induce cracking and failure - unless the reinforcement (longitudinal & links/stirrups) are detailed to cater for it for it. In the beam there will be additional longitudinal reinforcement in the top, bottom AND SIDES to cater for any torsion. The figures below may help you understand the "practical" on site concept of catering for torsion with properly detailed reinforcement. 1. What is torsion: 2. Below is a torsion "link" or "stirrup" shape – what it looks like: Below is a torsional situation: In Pakistan always include a bar in bottom (shown red) of cantilever balcony ALSO, same shape as top bar, but bend up and around towards to top of beam/slab, to allow for load reversal during an earthquake! When I built my house in Pakistan, the steel fixer managed to bent one bar thus for the balcony - the workers in Pakistan (where material is expensive and labour is cheap) are very talented, but do not have the technical know-how sadly..... that's the challenge for the likes of you the structural engineer to disseminate down! Keep an eye on this forum. There is Pakistan RC Building Reinforcement Detailing Manual in preparation and will include this torsion detail now that it has come up here! This Manual be an advisory document but will assist both "young" structural engineers and reinforcement detailers (draftsmen) to understand reinforcement detailing and fixing! Passing down knowledge – think of it as “zakat”, then it gets easier then to give and share!
    4 points
  33. I have observed in the past 6-month that there is a real need to bring together and promote the profession of Structural Engineering in Pakistan. To start the process we are working on a platform PASE - Pakistan Association of Structural Engineers (www.pase.pk). This is a voluntary setup and the aim is to promote practical aspects of structural design and construction in Pakistan. We have started with a simple website, which s in early stages of development (baby steps). Interesting to note that the idea came about after a post in SEPAKISTAN. Whilst nothing is easy in Pakistan, (which makes the challenge more exciting), we aim to make inroads and bring the "Pakistan Structural Engineering Profession" to International Standards by knowledge sharing on real practical design and construction topics - so you can be equipped to compete with the right skills in both national and Internationally against other consulting engineers. For forum and discussions, the PASE website directs the visitor to this SEPAKISTAN forum....which i think works great, so congratulations to Sepakistan team - great set up guys. First challenge is to compile a "Pakistan Building Reinforcement Detailing Manual", and make it available on the website in the next few months(only started on it after a post in this forum a few days ago, will take a month or two to compile). The aim is to equip the Pakistani Structural Engineers with knowledge and information on "practical structural design" - that is at par with international standards. A 2-day programme of training courses and lectures was in the planning for October 2020 in Islamabad - but this has had to be postponed for the time being due to Covid-19 social distancing and travel restrictions. This may now probably move to later in the year or even early in 2021 - we shall see. The website: pase.pk (If you have issue connecting then google "Pakistan Association of Structural Engineers pase.pk") - we certainly are not IT professionals. Any feedback via the "contact" tab of the website from structural engineers would be welcome. Kind Regards Simple Structures
    4 points
  34. Adding to above 3- Consult the seniors , discuss with them, have someone as mentor. 4-Go to the field and analyse the construction practices and compare them with the assumed ones for design. 5-Have ethics to accept your mistake if any reviewer mentions that and improve it. 6-Have ability to demonstrate what you have designed and how, do not follow the software blindly. 7-keep a diary, note problems and their proposed solutions, anything you need to learn or discuss and update it regularly, discuss problems with seniors/colleagues/teachers and analyse their solutions verify them and write in your diary. 8-Have some ethics and practice them, you must be able to analyse clearly even your own works and give an unbiased opinion. ... Regards
    4 points
  35. A good question. No base is fully pinned or fully fixed! 1. Engineering judgements are needed for base fixity, and experienced engineer use this judgement to analyse, design AND detail the column/base connection. 2. Fixed based require large footings - uneconomical in most instances; also, very difficult to get full fixity with a pad bases once the base rotates under moment. Only piled foundations with thick pile caps may offer anything close to full fixity! 3. Sometimes (not common) column flexural stiffness (EI/L, 4EI/L, 3EI/L etc) is used to derive "partial fixity" for analysis purposes. Majority of column/base connections are inherently partially fixed - but in analysis this is ignored. 4. I have analysed buildings with pinned bases, and then re-analysed the same frame for serviceability wind deflection assuming 10-20% fixity! All an engineering judgement call. 5. Pinned bases are used as a matter of course for most designs of concrete frames. There are exceptions to this rule: (a) for portal frame type crane buildings require fixed bases to limit lateral sway of frame (H/400), and make sure crane does not come off the crane rails; (b) Multi storey steel frames often use ‘fixed bases’ to limit frame sway at bottom, and design foundations accordingly. 6. In Pakistan I would always assume pinned bases for analysis purposes (although 10-20% fixity will always be inherent e connection of column/base, which helps but should not be assumed in analysis). Although this leads to higher moment on columns between first and second floors, and hence lower column needs to be sized the same as column above. 7. With a ‘fixed base’ frame the lowest column will attract greater damage to lowest column/beams during an earthquake. seismic regions. With ‘pinned base’ less damage to the column would occur and more to the beam ...beams are easier to strengthen later then columns! 8. My Conclusion: Use pinned bases in rc frame upto say 5-6 storey high for analysis purposes. Note design is half the story, make sure you sketch the critical reinforcement connection details for the steel detailer & fixer too. Hope this helps?
    4 points
  36. I believe the attached document would give you a little guidance to overall manual design process in load bearing residential homes Manual Design of Residential Homes.pdf
    4 points
  37. Dear @Saiful Islam Zaber Please see this link https://wiki.csiamerica.com/display/kb/Pattern+live-load+factor and the picture attached. Hope this will suffice your query.
    4 points
  38. So looks like today, I came across the same problem that is being discussed here. I had to provide a design criteria for a buried concrete pit. I will summarize my findings below for the benefit of everyone. This applies to structural members that are subjected to environmental exposure conditions or that are required to be liquid tight. 1) The first step is to calculate flexural demand in the walls of concrete pit/ water tank based on all possible conditions. For the case of buried concrete pit, it included, empty condition (no fluid in the pit) , operating condition (full of liquid), test condition (no backfill around the pit and it is full of liquid) etc. Buoyancy checks should also be performed. 2) Compare required flexural reinforcement against minimum reinforcement ratio = 0.006 per ACI 350, Table 7.12.2.1 & ACI 224 , Section 3.5 and provide whichever is the maximum. The ratios provided are basically temperature and shrinkage reinforcement ratios based on gross section so provide half of the reinforcement at each face. 3) Satisfy maximum crack width of water-retaining structure = 0.10 mm, ACI 224R-01 Table 4.1 based on the reinforcement already provided. If the reinforcement is inadequate, increase the reinforcement till this requirement is met. To meet this requirement, smaller bars should be used with close spacing. Now a few comments on the the excellent discussion above. @Khawaja Talha post above is applicable for all normal cases where there is a restraint to shrinkage and temperature movements only. If you have a condition like that, you need to provide 0.45% reinforcement ratio in your slabs. Example of a situation where this would be applicable will be a structure where movement or expansion joints haven't been provided at industry standard spacing. But if you want to meet liquid tight start with 0.6% as a minimum and work your way as suggested above. Other posts above explain the same things in a slightly different manner but all good. Thank you
    4 points
  39. I had tried to do the design of slabs and foundation in ETABS, it is very good and better than SAFE and easier. The drawback of designing the foundation in ETABS is that the full model work together and any change on the foundation sizes or thickness, it changes the stresses on the full system not only the foundation which force you to check the model more than one time to make sure your design is safe. Actually, this way leads to perfect design, but you will lose a lot of time in doing the alterations.
    4 points
  40. UmarMakhzumi

    Forum Update

    Dear All, The forum has been updated today with a lot new features. You can find the list of all the new improvements by visiting this website. Some highlights are: 1) Improved Search Features 2) Emoji Emoji support is now available in all editors. 3) Member profile cover photos are now clickable and can be viewed as a full image. Do check out mine as it is a very beautiful waterfall snap taken by yours truly. 😃 Do check out the link posted above for the complete list. One additional announcement that I would like to make is that with reference to last forum update post (read below), @Rana and @BAZ are forum Admins now. I think it was important to do as it brings more transparency for the forum and also helps spread the responsibility. The forum belongs to the members so it never made sense for one person to be Admin, As always, feedback is much appreciated. Thanks for taking the time out to read this update. Cheers!
    4 points
  41. In this case, you have to provide lap splices for column's lateral reinforcement please refer to ACI-318 2005 Item No. 7.10 page No. 87 for more details.... or check any latest version of ACI code. Thanks
    4 points
  42. Material behavior can be idealized as consisting of an 'elastic' domain and a 'plastic' domain. For almost 200 years, structural design has been based on an elastic theory which assumes that structures display a linear response throughout their loading history, ignoring the post-yielding stage of behavior. Current design practice for reinforced concrete structures is a curious blend of elastic analysis to compute forces and moments, plasticity theory to proportion cross-sections for the moment and axial, load, and empirical mumbo-jumbo to proportion members for shear. From the book "Design of Concrete Structures with Stress Fields" by A. Muttoni, J. Schwartz and B.Thurliman.
    4 points
  43. Very informative & impressive article, on a very very important topic of earthquake awareness, both technically as well as for the general knowledge of common people. And, that too in Urdu language for easy understanding of even non-technical people. Fully appreciated. Except a mistake (most probably typographic) of indicating Quetta in Zone 3 (Instead of Zone 4), the article is really superb, both from technical information as well from clarity points of view. Hopefully, it would help the readers understand how the earthquakes initiate, what are their negative effects & why their homes should be earthquake resistant. Regards.
    4 points
  44. AA. Here is the progress & update regarding the problem described in my original post, for the benefit of the interested SEFP users. 1. Automeshing was adopted in the original ETABS 9 model, with a maximum mesh size of 3 ft for the slabs & walls. The slabs included a raft slab as well, for tranferring the reactions to SAFE. 2. The model was being updated in steps, by refining the geometry along various edges of the slabs according to architectural plans at different floors. Model was running fine at all the steps. 3. The problem most probably occurred when the whole model (including the raft slab) was selected through 'Select all' command & automeshed again as stated in para-1 above. This action meshed the raft slab as well.The meshing of raft slab was haphazard in some regions, because of orientation of slab outer edges in different directions. 4. Following steps were taken in order to sort out the error:- a. Trials were made by automeshing the model at two different maximum mesh sizes of 4 ft & 2 ft (one at a time), but without any positive result. b. Third trial was made using 'default' meshing option ( meshing at grids & in Automesh options, by keeping the 'Further subdivide shell/wall in maximum element size of' option UNCHECKED. This trial helped in running the analysis completely, with 2 Warning Messages indicating presence of unconnected point & frame objects at the specified locations. Deletion of these extra objects removed the warning messages. c. Another trial indicated that thr problem is with the meshing of slabs, and not the walls. d. Trying to locate affected element # 22836 through the procedure advised by Rana, indicated that the relevant output table designates the slab & wall elements by the number types F### for slab elements & W### for wall elements. It followed that element numbers (like 22836 etc) are generated during analysis only. The same facts have been stated above by Saad Pervez. Thus, the affected element could not be traced once again. e. Keeping in view suggestion made by EngrJunaid, the affected model was opened in ETABS 2016. This time the affected element was located & found to be present in the automeshed Raft Slab, provided at the base level merely for exporting the Base Reactions to SAFE for further processing there. 5. Replacing the affected Automeshed Raft Slab, with a new unmeshed raft slab & remeshing the floor slabs & rc walls (after selecting the floor slabs & walls separately) at desired maximum mesh size finally solved the problem. In the end, thanks to all (especially @EngrJunaid) for contribution to reach the solution of subject problem, by suggesting various courses of action. Regards.
    4 points
  45. Top reinforcement is needed in isolated footing majorly for two reasons. 1) Due to negative pressure under some or whole part of footing. In part of footing where there is positive pressure, the footing is in complete contact with soil and tension is in bottom side of footing. But in part of footing where there is negative pressure the footing is no more in contact with ground. Either it is designed as it is (and reduced contact area is used for calculations) or it is made to become in contact with ground by help of over burden loads. In either case, the bending of footing is in such a way which causes tension of footing at top of foundation demanding top reinforcement in footing. Mostly top reinforcement is less than bottom reinforcement but for simplicity, if it is not affecting economy much same reinforcement can also be used for top and bottom. 2) Due to temperature and shrinkage control. Code says we can provide temperature based minimum steel either in one layer at center or at any face or we can divide total steel in two layers i.e. half at top face and half at bottom. According to Zahid Ahmad Siddique (Professor at UET Lahore) in his book concrete structures mentions it is better to provide temperature steel in two layers if thickness of footing increases 18 inch.
    4 points
  46. You can read more on https://en.wikipedia.org/wiki/Lotus_Temple . Here are a few pics. By Wiki-uk - Own work, CC BY-SA 3.0, https://commons.wikimedia.org/w/index.php?curid=18545739
    4 points
  47. UmarMakhzumi

    IRON RING

    I found this picture today while looking through archives. The Ring is visible on the little finger. Wear it all the time.
    4 points
  48. EngrJunaid

    Time Period

    In order to find the TIME PERIOD,seismic weight and Base shear used by ETABS in analysis, go to DISPLAY--->Show tables and display "Auto Siesmic Loads" in the Load Definitions under Model definition.... See the attached screen shots...
    4 points
  49. Great..!!! I m only familiar with Engr Junaid because he is my class fellow, with Umar Makhzumi Bhai, and With Waqar Saleem Brother. Today I came to know what is full form of BAZ And he has changed his looks also. By the way, I used to think Engr Uzair Sir quite younger but he is quite senior to us. Mostly people after getting senior does not use social media forums much.. especially for responding people.. Really nice to know the members. This forum is really great.!!
    4 points
  50. this book also useful NED research paper https://www.academia.edu/8983144/A_Practical_Guide_to_Nonlinear_Static_Analysis_of_Reinforced_Concrete_Buildings_with_Masonry_Infill_Walls
    4 points
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