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Syed Umair Haider

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  1. Like
    Syed Umair Haider got a reaction from Muhammad Zahid in Pin Support Or Fixed Support   
    As Umar said the real response would stand somewhere in between a pin and fixed base, similarly extent of column fixity at base depends upon the rotational flexibility of foundation that depends on foundation stiffness and soil stiffness both.
     
    In general it can be said that for rigid raft foundations,foundation supported on stiff piles or basement walls this rotational stiffness is high and one can confidently assume fixity of column at base.
     
    But in conditions of individual footing pads on deformable soil,foundation could have considerable rotational flexibility and consequently assumption of fixed column base could show a considerable variance with respect to distribution of column moments on bottom storey i.e in such case column moments could be concentrated at top end of bottom most column rather than at base.
     
    Therefore, It is recommended to model the rotational stiffness of foundation rather than assuming fixity at base in cases of individual column pads on deformable soil to represent quite realistic partial fixity available at column base.This methodology can be seen in "Seismic design of reinforced concrete and masonary building by M.J.N Priestley" (page 466).
  2. Like
    Syed Umair Haider got a reaction from shahidthanvi in Minimum Reinfocement Criteria For Crack Control   
    It seems that ACI is clear in its provision and usage of 0.6% rather than 0.2% reinforcing ratio for temperature & shrinkage crack/stress control could not be justified for general structures and is only applicable in cases where crack width is a design consideration.
     
    As ACI -350 table 7.12.2.1 recommends only for "structures subjected to environment or required to be liquid tight" a reinforcing ratio range from .003 to .006 for Grade 40 & .003 to .005 for Grade 60 steel as a function of length between movement joints from 20' to 40'.(see ACI 350 7.12.2.1).
    Thus the extreme ratios ".005" & ".006" represents the maximum temp & shrinkage reinf ratio for "structures subjected to environment or required to be liquid tight" where movement joint is not provided.This maximum ratio is mentioned in ACI 224 and is supposed to be followed when crack width is a design consideration for eg in structures as defined above.(note here structures subjected to environment could simply refer to outdoor structures in accordance with CIRIA technical note 107 section
     
    2) On the other hand as long as general structures is concerned (say building structures) crack width is not a design consideration in accordance with same code (ACI 318) and here provision of minimum reinf (0.18-0.2%) intends to limit shrinkage & temperature stresses rather than crack widths within tolerable limits (see ACI 318-11 7.12.1) which is claimed to be observed for many years satisfactorily controlled with these ratios.
     
    As long as design consideration of shrinkage & temperature effects is concerned in general structures, we include the effects of shrinkage & temp cracks indirectly in serviceability limit states as
     
    1, In the form of additional vertical deflection due to creep & shrinkage of structural members (see ACI 435R-95 table 2.1).
     
    2, Effective stiffness (see ACI 318-11 10.10.4) to be used in the analysis of lateral deflection accommodates the stiffness reduction due to shrinkage & temp cracking.
     
    3, ACI 318 also recommends to analyze for the severity of stresses arising from temp & shrinkage effects (see ACI 318-11 8.2.4 & 9.2.3)
     
    Hence, usage of 0.18-0.2% reinf and 0.3-0.6% reinf is justified for general and special structures respectively.
  3. Like
    Syed Umair Haider got a reaction from Ayesha in Issues in ETABS results   
    I think both ETABS and UBC 97 provisions are correct.
    As ETABS completely follows UBC97 therefore only UBC's provisions are elaborated below.
    UBC seismic design philosophy limits the minimum time period (in the form of maximum base shear ) to Ts (UBC response spectrum) which is equivalent to Cv/2.5Ca.
    For elaboration ,
    V=CV.I.W/R.T
    Substitutue T=Ts=Cv/2.5Ca in above eqn
    therefore 
    V=(CV.I.W/R)X(2.5Ca/Cv)
    i.e V=2.5Ca.I/w (the maximum base shear that ETABS and UBC97 uses)
    Therefore time period regardless of its shorter value below Ta cannot be lesser than Ts which is in contrast with UBC97 provisions that are inherent in ETABS.
    However, analysis indicating time period lesser than Ta indicates a over stiff/over design structure, therefore it is recommended to maintain a natural period closer to 0.1x (no of floors).
     
     
     
     
  4. Thanks
    Syed Umair Haider got a reaction from Omer Ahmed in Design For Shear And Torsion Using Etabs   
    Dear Zain,
     
    I don't know the origin of document,you have uploaded for calculating torsional constant,but the methodology given therein is incorrect.As "Tcr" and "Tu" given therein are indeed threshold torsional strength and ultimate torsional stresses respectively, and are both design properties  not analysis properties. (See ACI 318-11 section 11.5.1).
     
    Whereas the torsional constant, ETABS asks in "analysis property modification factors" is simply the torsional moment of inertia (J) used to determine torsional stiffness of a member (JG/L) i.e something else.
     
    As long as its value is concerned,then in building structures it is a general practice to use a negligible value like .001 to nullify beam's torsional stiffness.In this way, the torsional stresses (if arising due to compatibility of deformation i.e compatibility torsion ) are transferred via alternate load path (i.e redistribution of torsional moments occurred), considering that beam is unable to provide torsional restraint and in other condition if torsional stresses in beam is required to satisfy equilibrium of structure (where redistribution is not possible) then torsional stresses in beams remains independent of whatever value of "J" you have selected as equilibrium equations are necessarily satisfied independent of stiffness as "Compatibility is optional and equilibrium is essential".
     
    This approach of minimization of "J" economize beam sizes that arise from stringent combined shear and torsion requirement of building codes,but consequently beam sections designed in this way will start developing internal horizontal cracks (hairline cracks not affecting functionality of structure) due to torsional stresses and their torsional strength will continuously degrade till the design condition is achieved i.e negligible torsional strength of beam.But as the structure is designed to be stable without torsional stiffness of beam so it remain stable after this condition is achieved.However, the beam member itself cracks that doesn't affect the functionality of structure in any way.
     
    A very descriptive and clarifying description is available in "Reinforced concrete design by Arthur Nilson".
     
    As long as authentication of this approach is concerned then it is allowed by building codes as,
     
    1, ACI-318-11 section 11.5.2.1 & 11.5.2.2.
     
    2, UBC97 section 1911.6.2.1 & 1911.6.2.2
     
    3, BS 8110-1 1997 section 3.4.5.13
     
    Keeping in view above mentioned, it is a general practice to nullify torsional constant of beams in building structures and it is not required to use any iterative process to derive torsional constant of each beam section that is indeed not practical as there will be thousands of beam span in large structures.
  5. Like
    Syed Umair Haider got a reaction from Padika in Minimum Reinfocement Criteria For Crack Control   
    It seems that ACI is clear in its provision and usage of 0.6% rather than 0.2% reinforcing ratio for temperature & shrinkage crack/stress control could not be justified for general structures and is only applicable in cases where crack width is a design consideration.
     
    As ACI -350 table 7.12.2.1 recommends only for "structures subjected to environment or required to be liquid tight" a reinforcing ratio range from .003 to .006 for Grade 40 & .003 to .005 for Grade 60 steel as a function of length between movement joints from 20' to 40'.(see ACI 350 7.12.2.1).
    Thus the extreme ratios ".005" & ".006" represents the maximum temp & shrinkage reinf ratio for "structures subjected to environment or required to be liquid tight" where movement joint is not provided.This maximum ratio is mentioned in ACI 224 and is supposed to be followed when crack width is a design consideration for eg in structures as defined above.(note here structures subjected to environment could simply refer to outdoor structures in accordance with CIRIA technical note 107 section
     
    2) On the other hand as long as general structures is concerned (say building structures) crack width is not a design consideration in accordance with same code (ACI 318) and here provision of minimum reinf (0.18-0.2%) intends to limit shrinkage & temperature stresses rather than crack widths within tolerable limits (see ACI 318-11 7.12.1) which is claimed to be observed for many years satisfactorily controlled with these ratios.
     
    As long as design consideration of shrinkage & temperature effects is concerned in general structures, we include the effects of shrinkage & temp cracks indirectly in serviceability limit states as
     
    1, In the form of additional vertical deflection due to creep & shrinkage of structural members (see ACI 435R-95 table 2.1).
     
    2, Effective stiffness (see ACI 318-11 10.10.4) to be used in the analysis of lateral deflection accommodates the stiffness reduction due to shrinkage & temp cracking.
     
    3, ACI 318 also recommends to analyze for the severity of stresses arising from temp & shrinkage effects (see ACI 318-11 8.2.4 & 9.2.3)
     
    Hence, usage of 0.18-0.2% reinf and 0.3-0.6% reinf is justified for general and special structures respectively.
  6. Like
    Syed Umair Haider got a reaction from Ayesha in Minimum Reinfocement Criteria For Crack Control   
    It seems that ACI is clear in its provision and usage of 0.6% rather than 0.2% reinforcing ratio for temperature & shrinkage crack/stress control could not be justified for general structures and is only applicable in cases where crack width is a design consideration.
     
    As ACI -350 table 7.12.2.1 recommends only for "structures subjected to environment or required to be liquid tight" a reinforcing ratio range from .003 to .006 for Grade 40 & .003 to .005 for Grade 60 steel as a function of length between movement joints from 20' to 40'.(see ACI 350 7.12.2.1).
    Thus the extreme ratios ".005" & ".006" represents the maximum temp & shrinkage reinf ratio for "structures subjected to environment or required to be liquid tight" where movement joint is not provided.This maximum ratio is mentioned in ACI 224 and is supposed to be followed when crack width is a design consideration for eg in structures as defined above.(note here structures subjected to environment could simply refer to outdoor structures in accordance with CIRIA technical note 107 section
     
    2) On the other hand as long as general structures is concerned (say building structures) crack width is not a design consideration in accordance with same code (ACI 318) and here provision of minimum reinf (0.18-0.2%) intends to limit shrinkage & temperature stresses rather than crack widths within tolerable limits (see ACI 318-11 7.12.1) which is claimed to be observed for many years satisfactorily controlled with these ratios.
     
    As long as design consideration of shrinkage & temperature effects is concerned in general structures, we include the effects of shrinkage & temp cracks indirectly in serviceability limit states as
     
    1, In the form of additional vertical deflection due to creep & shrinkage of structural members (see ACI 435R-95 table 2.1).
     
    2, Effective stiffness (see ACI 318-11 10.10.4) to be used in the analysis of lateral deflection accommodates the stiffness reduction due to shrinkage & temp cracking.
     
    3, ACI 318 also recommends to analyze for the severity of stresses arising from temp & shrinkage effects (see ACI 318-11 8.2.4 & 9.2.3)
     
    Hence, usage of 0.18-0.2% reinf and 0.3-0.6% reinf is justified for general and special structures respectively.
  7. Like
    Syed Umair Haider got a reaction from Hafsa Azmat in Hinge or Fixed Support of Columns for Pile Loads computation   
    Dear All,
    Kindly share your experience regarding type of support assignment (hinge or fixed) in super structure models for computation of pile loads in case of,
    a, Columns supported in isolated pile caps.
    b, Columns supported on piled mats.
    Emphasizing on following parameters,
    1, Difference in Pile loads extracted between both cases.
    2, Difference in super structure column design between both cases.
    3, Difference in lateral stability of super structure between both cases.
     
     
     
  8. Thanks
    Syed Umair Haider got a reaction from Vamshi Prasad in Design For Shear And Torsion Using Etabs   
    Dear Zain,
     
    I don't know the origin of document,you have uploaded for calculating torsional constant,but the methodology given therein is incorrect.As "Tcr" and "Tu" given therein are indeed threshold torsional strength and ultimate torsional stresses respectively, and are both design properties  not analysis properties. (See ACI 318-11 section 11.5.1).
     
    Whereas the torsional constant, ETABS asks in "analysis property modification factors" is simply the torsional moment of inertia (J) used to determine torsional stiffness of a member (JG/L) i.e something else.
     
    As long as its value is concerned,then in building structures it is a general practice to use a negligible value like .001 to nullify beam's torsional stiffness.In this way, the torsional stresses (if arising due to compatibility of deformation i.e compatibility torsion ) are transferred via alternate load path (i.e redistribution of torsional moments occurred), considering that beam is unable to provide torsional restraint and in other condition if torsional stresses in beam is required to satisfy equilibrium of structure (where redistribution is not possible) then torsional stresses in beams remains independent of whatever value of "J" you have selected as equilibrium equations are necessarily satisfied independent of stiffness as "Compatibility is optional and equilibrium is essential".
     
    This approach of minimization of "J" economize beam sizes that arise from stringent combined shear and torsion requirement of building codes,but consequently beam sections designed in this way will start developing internal horizontal cracks (hairline cracks not affecting functionality of structure) due to torsional stresses and their torsional strength will continuously degrade till the design condition is achieved i.e negligible torsional strength of beam.But as the structure is designed to be stable without torsional stiffness of beam so it remain stable after this condition is achieved.However, the beam member itself cracks that doesn't affect the functionality of structure in any way.
     
    A very descriptive and clarifying description is available in "Reinforced concrete design by Arthur Nilson".
     
    As long as authentication of this approach is concerned then it is allowed by building codes as,
     
    1, ACI-318-11 section 11.5.2.1 & 11.5.2.2.
     
    2, UBC97 section 1911.6.2.1 & 1911.6.2.2
     
    3, BS 8110-1 1997 section 3.4.5.13
     
    Keeping in view above mentioned, it is a general practice to nullify torsional constant of beams in building structures and it is not required to use any iterative process to derive torsional constant of each beam section that is indeed not practical as there will be thousands of beam span in large structures.
  9. Thanks
    Syed Umair Haider got a reaction from Sunita in Design For Shear And Torsion Using Etabs   
    Dear Zain,
     
    I don't know the origin of document,you have uploaded for calculating torsional constant,but the methodology given therein is incorrect.As "Tcr" and "Tu" given therein are indeed threshold torsional strength and ultimate torsional stresses respectively, and are both design properties  not analysis properties. (See ACI 318-11 section 11.5.1).
     
    Whereas the torsional constant, ETABS asks in "analysis property modification factors" is simply the torsional moment of inertia (J) used to determine torsional stiffness of a member (JG/L) i.e something else.
     
    As long as its value is concerned,then in building structures it is a general practice to use a negligible value like .001 to nullify beam's torsional stiffness.In this way, the torsional stresses (if arising due to compatibility of deformation i.e compatibility torsion ) are transferred via alternate load path (i.e redistribution of torsional moments occurred), considering that beam is unable to provide torsional restraint and in other condition if torsional stresses in beam is required to satisfy equilibrium of structure (where redistribution is not possible) then torsional stresses in beams remains independent of whatever value of "J" you have selected as equilibrium equations are necessarily satisfied independent of stiffness as "Compatibility is optional and equilibrium is essential".
     
    This approach of minimization of "J" economize beam sizes that arise from stringent combined shear and torsion requirement of building codes,but consequently beam sections designed in this way will start developing internal horizontal cracks (hairline cracks not affecting functionality of structure) due to torsional stresses and their torsional strength will continuously degrade till the design condition is achieved i.e negligible torsional strength of beam.But as the structure is designed to be stable without torsional stiffness of beam so it remain stable after this condition is achieved.However, the beam member itself cracks that doesn't affect the functionality of structure in any way.
     
    A very descriptive and clarifying description is available in "Reinforced concrete design by Arthur Nilson".
     
    As long as authentication of this approach is concerned then it is allowed by building codes as,
     
    1, ACI-318-11 section 11.5.2.1 & 11.5.2.2.
     
    2, UBC97 section 1911.6.2.1 & 1911.6.2.2
     
    3, BS 8110-1 1997 section 3.4.5.13
     
    Keeping in view above mentioned, it is a general practice to nullify torsional constant of beams in building structures and it is not required to use any iterative process to derive torsional constant of each beam section that is indeed not practical as there will be thousands of beam span in large structures.
  10. Like
    Syed Umair Haider got a reaction from israr_sari in Property Modifiers For Retaining Walls   
    Asad,
    It is not necessarily required to extend the seismic analysis's storey range till basement levels as for buildings with several below grade levels supported by basement walls, two stage static analysis procedure is used (ASCE 7-10 Section 12 & UBC97 section 1630.4.2) that consists in distribution of building in flexible upper portion (above basement levels) and rigid lower portion (basement levels), provided the lower portion have a stiffness minimum 10 times greater than upper and time period of whole structure should not exceed 1.1 times of flexible upper portion's period while it is considered as a separate structure.
    You can simply check these limitations as,
    1, by computing stiffness ratio (EI/L ratios of basement walls + LFRS in rigid lower portion) to the (EI/L ratios of LFRS in flexible upper portion)
    2, computing time period of whole structure (Eigen vector) and computing time period of upper portion alone modeled without basement levels.
    Having satisfied these, seismic analysis is required to be performed till base of upper portion only & rigid lower portion is required to design only for seismic forces transmitted at the base of flexible upper portion modified by the factor Rupper/Rlower.
    In ETABS you have to define "ground level" as bottom storey in analysis storey range and seismic shear imparted on ground level will be automatically transmitted to the levels below through diaphragm action.It will be just required to compute "R" value for lower portion considering it separate and to modify seismic load case's scale factor by Ru/Rl for the design of below grade structure.
    In this way the maximum seismic shear will be acting at the ground level not at B4, that will reduce the magnitude of force and could be beneficial in mentioned below grade serviceability issues particularly drift will be considerably reduced (also compute drift using user defined time period obtained from eigen vector analysis see UBC Section 1630.10.3).
    As long as below grade torsion is concerned, it is just required to satisfy that Ax (UBC97 Eqn 30-16) should not exceed 3 and required to be noted that amplification of diaphragm eccentricity is of no meaning there since seismic forces are imposed from upper portion and are not calculated & applied separately.
    Secondly, load combinations should be inclusive of minimum seismic vertical effects and dynamic load combinations.
  11. Like
    Syed Umair Haider got a reaction from UmarMakhzumi in Issues in ETABS results   
    Dear Waseem
    Your concern is valid but the lower limit given in FEMA example doesn't make sense to me as in a general the usage of time period in seismic analysis is not more than the evaluation of ground acceleration that will be imparted in structure in accordance with specific structural characteristics.
    Similarly if the specific structural characteristics of a building reveals a time period lesser than generalized approximate time period , then it doesn't make sense to skip the accurate time period and to use approximate one when code itself defines approximate time period as a basis to start analysis for actual T.
    More interesting is to note that indicated FEMA example is based on 2009 NEHRP seismic provisions , whereas in a separate document issued by NEHRP afterwards (named expanded seismic commentary to ACE-10) it is recommended to use "Tcomp" if "Tcomp < Ta".
    Moreover, i didn't find any lower bound on time period in UBC97 and even in ASCE 7 which is based on same ELF procedure as given in FEMA example.
    Therefore, it seems that indicated provisions in FEMA example are overlooked that they have fixed later and usage of Tcomputed if "Tcomputed < Tapproximate" seems valid.
     
     
     
  12. Like
    Syed Umair Haider got a reaction from Ayesha in Issues in ETABS results   
    Dear Waseem
    Your concern is valid but the lower limit given in FEMA example doesn't make sense to me as in a general the usage of time period in seismic analysis is not more than the evaluation of ground acceleration that will be imparted in structure in accordance with specific structural characteristics.
    Similarly if the specific structural characteristics of a building reveals a time period lesser than generalized approximate time period , then it doesn't make sense to skip the accurate time period and to use approximate one when code itself defines approximate time period as a basis to start analysis for actual T.
    More interesting is to note that indicated FEMA example is based on 2009 NEHRP seismic provisions , whereas in a separate document issued by NEHRP afterwards (named expanded seismic commentary to ACE-10) it is recommended to use "Tcomp" if "Tcomp < Ta".
    Moreover, i didn't find any lower bound on time period in UBC97 and even in ASCE 7 which is based on same ELF procedure as given in FEMA example.
    Therefore, it seems that indicated provisions in FEMA example are overlooked that they have fixed later and usage of Tcomputed if "Tcomputed < Tapproximate" seems valid.
     
     
     
  13. Like
    Syed Umair Haider got a reaction from Badar (BAZ) in Issues in ETABS results   
    Dear Waseem
    Your concern is valid but the lower limit given in FEMA example doesn't make sense to me as in a general the usage of time period in seismic analysis is not more than the evaluation of ground acceleration that will be imparted in structure in accordance with specific structural characteristics.
    Similarly if the specific structural characteristics of a building reveals a time period lesser than generalized approximate time period , then it doesn't make sense to skip the accurate time period and to use approximate one when code itself defines approximate time period as a basis to start analysis for actual T.
    More interesting is to note that indicated FEMA example is based on 2009 NEHRP seismic provisions , whereas in a separate document issued by NEHRP afterwards (named expanded seismic commentary to ACE-10) it is recommended to use "Tcomp" if "Tcomp < Ta".
    Moreover, i didn't find any lower bound on time period in UBC97 and even in ASCE 7 which is based on same ELF procedure as given in FEMA example.
    Therefore, it seems that indicated provisions in FEMA example are overlooked that they have fixed later and usage of Tcomputed if "Tcomputed < Tapproximate" seems valid.
     
     
     
  14. Like
    Syed Umair Haider got a reaction from UmarMakhzumi in Issues in ETABS results   
    I think both ETABS and UBC 97 provisions are correct.
    As ETABS completely follows UBC97 therefore only UBC's provisions are elaborated below.
    UBC seismic design philosophy limits the minimum time period (in the form of maximum base shear ) to Ts (UBC response spectrum) which is equivalent to Cv/2.5Ca.
    For elaboration ,
    V=CV.I.W/R.T
    Substitutue T=Ts=Cv/2.5Ca in above eqn
    therefore 
    V=(CV.I.W/R)X(2.5Ca/Cv)
    i.e V=2.5Ca.I/w (the maximum base shear that ETABS and UBC97 uses)
    Therefore time period regardless of its shorter value below Ta cannot be lesser than Ts which is in contrast with UBC97 provisions that are inherent in ETABS.
    However, analysis indicating time period lesser than Ta indicates a over stiff/over design structure, therefore it is recommended to maintain a natural period closer to 0.1x (no of floors).
     
     
     
     
  15. Like
    Syed Umair Haider got a reaction from Badar (BAZ) in Issues in ETABS results   
    I think both ETABS and UBC 97 provisions are correct.
    As ETABS completely follows UBC97 therefore only UBC's provisions are elaborated below.
    UBC seismic design philosophy limits the minimum time period (in the form of maximum base shear ) to Ts (UBC response spectrum) which is equivalent to Cv/2.5Ca.
    For elaboration ,
    V=CV.I.W/R.T
    Substitutue T=Ts=Cv/2.5Ca in above eqn
    therefore 
    V=(CV.I.W/R)X(2.5Ca/Cv)
    i.e V=2.5Ca.I/w (the maximum base shear that ETABS and UBC97 uses)
    Therefore time period regardless of its shorter value below Ta cannot be lesser than Ts which is in contrast with UBC97 provisions that are inherent in ETABS.
    However, analysis indicating time period lesser than Ta indicates a over stiff/over design structure, therefore it is recommended to maintain a natural period closer to 0.1x (no of floors).
     
     
     
     
  16. Like
    Syed Umair Haider got a reaction from Waqas Haider in Deflection Calculations : Manual vs Etabs   
    Dear waqas,
    Calculation of Ie for live load only doesn't make sense as live load will always act on the section after the application of dead loads and will act simultaneously with dead loads.
    Therefore , Ie must be calculated for possible combinations that could be 1.0D,1.0D+0.5L & 1.0D+1.0L as per ACI-435.
    Detailed calculations of deflection for 1-way NP flexural members as per ACI 435 is shown in attachment.
     
     
     
     


  17. Like
    Syed Umair Haider got a reaction from UmarMakhzumi in Choice between rectangular and Tee Beams   
    W salam,
    In monolithic construction , T or rec beam is not a matter of choice of designer but its indicated by analysis that if depth of compression block lies below flange (flange = slab thk in beam-slab system) then the compressive force in concrete is balanced by a certain width of flange + total width of web, otherwise only web balance compressive force i.e example of rec section carrying slab load in monolithic construction.
    Example of rec beam sections also exists in some specific cases of construction for eg RC beams supporting metal deck system,hollow core slab panels and precast construction system.
    ETABS checks the eqn "a < or > ds" for maximum analyzed Mu-bott and design positive R/F accordingly.
     
     
  18. Like
    Syed Umair Haider got a reaction from Waqas Haider in Choice between rectangular and Tee Beams   
    W salam,
    In monolithic construction , T or rec beam is not a matter of choice of designer but its indicated by analysis that if depth of compression block lies below flange (flange = slab thk in beam-slab system) then the compressive force in concrete is balanced by a certain width of flange + total width of web, otherwise only web balance compressive force i.e example of rec section carrying slab load in monolithic construction.
    Example of rec beam sections also exists in some specific cases of construction for eg RC beams supporting metal deck system,hollow core slab panels and precast construction system.
    ETABS checks the eqn "a < or > ds" for maximum analyzed Mu-bott and design positive R/F accordingly.
     
     
  19. Like
    Syed Umair Haider reacted to Badar (BAZ) in Deflection Calculations : Manual vs Etabs   
    The modifier -0.35- is not meant to be used for calculating deflections of flexural members for gravity loads; It is to be used for calculating lateral deflections of the frame.
    You can make multiple models on Etabs to perform the calculation.
    It is better to use excel sheet. I do not perform these calculations on ETABS.
    Safe is another option to perform these calculations.
  20. Like
    Syed Umair Haider got a reaction from Bilal Shah in Ritz Analysis Problem   
    Auto meshing doesn't ensure adequate connectivity between member to member and is therefore recommended for horizontal area elements enclosed by line elements only (floors), where there is no structural connection between floor and any element in between the panel.
     
    In case of vertical elements, connectivity between vertical and horizontal elements is of due importance and is better to be achieved through manual meshing.
     
    In case of auto meshing as you indicated, change in size of auto mesh could solve the problem as its possible that connections inadequate (nodes not coinciding) on 1m element size can be adequate for 1.2m size (nodes start coinciding) and so on.
     
    For p-delta,a possibility exists that due to any meshing error some connection is modelled with inadequate lateral stiffness i.e when program try to impose lateral deflection due to seismic loads,modes start yielding frequency below shift.
     
    If you are interested in studying the problem,then easy approach is to check each mode shape and investigate the member that is going in unrealistically large displacement. Solving this member's connectivity inadequacy will solve your problem.
  21. Like
    Syed Umair Haider got a reaction from EngrJunaid in Design For Shear And Torsion Using Etabs   
    Dear Zain,
     
    I don't know the origin of document,you have uploaded for calculating torsional constant,but the methodology given therein is incorrect.As "Tcr" and "Tu" given therein are indeed threshold torsional strength and ultimate torsional stresses respectively, and are both design properties  not analysis properties. (See ACI 318-11 section 11.5.1).
     
    Whereas the torsional constant, ETABS asks in "analysis property modification factors" is simply the torsional moment of inertia (J) used to determine torsional stiffness of a member (JG/L) i.e something else.
     
    As long as its value is concerned,then in building structures it is a general practice to use a negligible value like .001 to nullify beam's torsional stiffness.In this way, the torsional stresses (if arising due to compatibility of deformation i.e compatibility torsion ) are transferred via alternate load path (i.e redistribution of torsional moments occurred), considering that beam is unable to provide torsional restraint and in other condition if torsional stresses in beam is required to satisfy equilibrium of structure (where redistribution is not possible) then torsional stresses in beams remains independent of whatever value of "J" you have selected as equilibrium equations are necessarily satisfied independent of stiffness as "Compatibility is optional and equilibrium is essential".
     
    This approach of minimization of "J" economize beam sizes that arise from stringent combined shear and torsion requirement of building codes,but consequently beam sections designed in this way will start developing internal horizontal cracks (hairline cracks not affecting functionality of structure) due to torsional stresses and their torsional strength will continuously degrade till the design condition is achieved i.e negligible torsional strength of beam.But as the structure is designed to be stable without torsional stiffness of beam so it remain stable after this condition is achieved.However, the beam member itself cracks that doesn't affect the functionality of structure in any way.
     
    A very descriptive and clarifying description is available in "Reinforced concrete design by Arthur Nilson".
     
    As long as authentication of this approach is concerned then it is allowed by building codes as,
     
    1, ACI-318-11 section 11.5.2.1 & 11.5.2.2.
     
    2, UBC97 section 1911.6.2.1 & 1911.6.2.2
     
    3, BS 8110-1 1997 section 3.4.5.13
     
    Keeping in view above mentioned, it is a general practice to nullify torsional constant of beams in building structures and it is not required to use any iterative process to derive torsional constant of each beam section that is indeed not practical as there will be thousands of beam span in large structures.
  22. Like
    Syed Umair Haider got a reaction from Badar (BAZ) in Aci 21.1.1 Energy Dissipation Confusion   
    W salam,
     
    These concepts need some fundamental theory of development of seismic analysis procedures (static procedures) in codes as briefed below,
     
    When engineers decided to go for an earth quake resistant design,then they initially proposed to assign a horizontal load of "0.1 x Weight of structure"  to cater for seismic forces.With the passage of time several geo-technical and site specific response characteristics were included in analysis for evaluation of seismic forces and structural members were designed to resist these forces in their elastic range i.e to not yield under these forces.
     
    Structures designed accordingly surprised engineers, as they were observed to show little tolerable non structural damages in seismic events considerably greater then those considered in evaluation of seismic forces.This leads to the development of concept of energy dissipation and over strength factor i.e under cyclic seismic loading structures have the ability of resistance beyond the elastic range of stresses in members (after yield), in proportion to their ductility.Since then started consideration of this structural over strength characteristics that consists in reduction of design seismic forces in accordance with their energy dissipation characteristics or mathematically reduction of base shear by division with over strength factor.
     
    For ex in accordance with UBC97, if on a structure the actual seismic force i.e Cv.I/W=550 & over strength or ductility factor is R=5.5 then Seismic shear will be 550/5.5 = 100, then structural members will be designed to remain elastic or not yield under the lateral force of 100, whereas they will dissipate the remaining 450 in inelastic range or in terms of energy it can be said that this structure is able to dissipate 450/550x100 = 81% seismic forces through its ductility and is required to design elastic only for 19% of actual seismic forces.
     
    In the lights of above these clauses could be defined as follows,
     
    ") 21.1.1 says, ........................................For which, design forces , related to earth quack forces, have been determined on the bases of ENERGY DISSIPATION IN NONLINEAR RANGE OF RESPONSE".
     
     For every structure,seismic forces are evaluated in accordance with corresponding over strength factor that indicates the extent of probable energy dissipation.
     
    2) Commentary of R 21.1.1 says, 
    The integrity of the structure in the inelastic range of response should be maintained because the design earth quack forces, defined in documents such as ASCE/SEI 7, the IBC,  the UBC and NEHRP provisions are considered less than those corresponding to linear response at the anticipated earthquack intensity. 
     
    As defined above, structures are designed for seismic forces that are reduced by over strength factor however actual fores are times greater than that, therefore code requires that when structure is subjected to actual seismic forces(plastic state),then although structural damages in members are tolerable but integrity of structural members should necessarily be maintained so that structure will not collapse.This condition is another form of philosophy of safety in code under seismic events that says "Under major earth quake,structure should be designed to have structural & non structural damages but should not collapse".
  23. Like
    Syed Umair Haider got a reaction from UmarMakhzumi in Aci 21.1.1 Energy Dissipation Confusion   
    W salam,
     
    These concepts need some fundamental theory of development of seismic analysis procedures (static procedures) in codes as briefed below,
     
    When engineers decided to go for an earth quake resistant design,then they initially proposed to assign a horizontal load of "0.1 x Weight of structure"  to cater for seismic forces.With the passage of time several geo-technical and site specific response characteristics were included in analysis for evaluation of seismic forces and structural members were designed to resist these forces in their elastic range i.e to not yield under these forces.
     
    Structures designed accordingly surprised engineers, as they were observed to show little tolerable non structural damages in seismic events considerably greater then those considered in evaluation of seismic forces.This leads to the development of concept of energy dissipation and over strength factor i.e under cyclic seismic loading structures have the ability of resistance beyond the elastic range of stresses in members (after yield), in proportion to their ductility.Since then started consideration of this structural over strength characteristics that consists in reduction of design seismic forces in accordance with their energy dissipation characteristics or mathematically reduction of base shear by division with over strength factor.
     
    For ex in accordance with UBC97, if on a structure the actual seismic force i.e Cv.I/W=550 & over strength or ductility factor is R=5.5 then Seismic shear will be 550/5.5 = 100, then structural members will be designed to remain elastic or not yield under the lateral force of 100, whereas they will dissipate the remaining 450 in inelastic range or in terms of energy it can be said that this structure is able to dissipate 450/550x100 = 81% seismic forces through its ductility and is required to design elastic only for 19% of actual seismic forces.
     
    In the lights of above these clauses could be defined as follows,
     
    ") 21.1.1 says, ........................................For which, design forces , related to earth quack forces, have been determined on the bases of ENERGY DISSIPATION IN NONLINEAR RANGE OF RESPONSE".
     
     For every structure,seismic forces are evaluated in accordance with corresponding over strength factor that indicates the extent of probable energy dissipation.
     
    2) Commentary of R 21.1.1 says, 
    The integrity of the structure in the inelastic range of response should be maintained because the design earth quack forces, defined in documents such as ASCE/SEI 7, the IBC,  the UBC and NEHRP provisions are considered less than those corresponding to linear response at the anticipated earthquack intensity. 
     
    As defined above, structures are designed for seismic forces that are reduced by over strength factor however actual fores are times greater than that, therefore code requires that when structure is subjected to actual seismic forces(plastic state),then although structural damages in members are tolerable but integrity of structural members should necessarily be maintained so that structure will not collapse.This condition is another form of philosophy of safety in code under seismic events that says "Under major earth quake,structure should be designed to have structural & non structural damages but should not collapse".
  24. Like
    Syed Umair Haider got a reaction from UmarMakhzumi in Ritz Analysis Problem   
    Auto meshing doesn't ensure adequate connectivity between member to member and is therefore recommended for horizontal area elements enclosed by line elements only (floors), where there is no structural connection between floor and any element in between the panel.
     
    In case of vertical elements, connectivity between vertical and horizontal elements is of due importance and is better to be achieved through manual meshing.
     
    In case of auto meshing as you indicated, change in size of auto mesh could solve the problem as its possible that connections inadequate (nodes not coinciding) on 1m element size can be adequate for 1.2m size (nodes start coinciding) and so on.
     
    For p-delta,a possibility exists that due to any meshing error some connection is modelled with inadequate lateral stiffness i.e when program try to impose lateral deflection due to seismic loads,modes start yielding frequency below shift.
     
    If you are interested in studying the problem,then easy approach is to check each mode shape and investigate the member that is going in unrealistically large displacement. Solving this member's connectivity inadequacy will solve your problem.
  25. Like
    Syed Umair Haider got a reaction from mhdhamood in Ritz Analysis Problem   
    Auto meshing doesn't ensure adequate connectivity between member to member and is therefore recommended for horizontal area elements enclosed by line elements only (floors), where there is no structural connection between floor and any element in between the panel.
     
    In case of vertical elements, connectivity between vertical and horizontal elements is of due importance and is better to be achieved through manual meshing.
     
    In case of auto meshing as you indicated, change in size of auto mesh could solve the problem as its possible that connections inadequate (nodes not coinciding) on 1m element size can be adequate for 1.2m size (nodes start coinciding) and so on.
     
    For p-delta,a possibility exists that due to any meshing error some connection is modelled with inadequate lateral stiffness i.e when program try to impose lateral deflection due to seismic loads,modes start yielding frequency below shift.
     
    If you are interested in studying the problem,then easy approach is to check each mode shape and investigate the member that is going in unrealistically large displacement. Solving this member's connectivity inadequacy will solve your problem.
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