UmarMakhzumi

AOA everyone. Recently I've been very confused with development length and how its provided in different components. I have a few conceptual confusions about it and then a few practical and design aspects of it. I'd really appreciate help on these.
Do we provide/check development length for simply supported beams (i.e. check it past the max moment?) Where do we need to check for development length? (in my understanding, we need to check it almost everywhere, where you need to ensure full moment capacity) What is a "critical section" when we talk about development length? (refer to the attached picture) I think its the point beyond which we need to ensure development At a wall-slab joint where both the slab and wall are 8" thick, if #6 bars at 8" c/c need to be developed from the slab into the wall, I found that even with the 90 degree hook, the development length is at least 14". How do you satisfy development length in such a case? For hooked development length ldh, is it the length up to the hook or is the length that includes the length of the hook? For example for a 90 degree hook, would it be 12db+the length beyond the critical section or just the length beyond the critical section up to the hook? I've seen in a few places on the internet that people include the 12db into the development length.  This stuff has been eating up gray matter from my brain for a few days now. Any help is appreciated.
Thanks!
 

There are typical details of reinforcement, which can be found in various detailing manuals, that the design engineers follow around the world. They do not normally check development length for typical/usual member sizes with usual/ conventional loadings, instead they follow those reinforcement details.
 
The critical section is the section with maximum tensile stress. For most beams or flexural members, it is the face of column or support.
 
For this particular case, you can treat the wall-end as simply supported. But, there are other ways to develop the reinforcement such as mechanical anchorages (headed bar) or welding with end-plates.
 
As you can see in your attached diagram, the length of hook is not considered by American codes. Some codes do consider them explicitly. For American codes, any length beyond 12db does not contribute.

I guess you are talking about modal super-position, but that is not my question.
 
Yes and according to SK Gosh (http://skghoshassociates.com/SKGAblog/viewpost.php?id=5), we need to amplify dynamic torsion because;
"....  accidental torsion is not determined as part of the dynamic analysis, but as the result of a separate static load applied at an eccentricity. The only way we can use the exception [to ignore amplification in dynamic] is to incorporate the accidental torsion effects into the building model itself by defining a floor mass distribution that is not uniform so that the center of mass has a 5% offset from the centroid of the floor area."
Agreed. Now, the next step is to how to do it practically. We are not just talking about the concept but to actually do it.
Now, see CSi ETABS Wiki (https://wiki.csiamerica.com/display/etabs/Accidental+eccentricity)
1. In the first method, accidental torsion is included in dynamic analysis by actually shifting CM as SK Gosh suggests that would change the dynamic properties, natural characteristics and stiffness matrices for each eccentricity and so we do not need to amplify further, because accidental torsion has been calculated through dynamic analysis.
Con: The main drawback is that as the properties of model change for each eccentricity, hence 4 separate eigenvalues analyses must be performed for each eccentricity and then finding a way to envelope the maximum response of these 4, which is not possible in majority of software. What is the simple solution? See point #3.
2. Second way is to model a static torsional load [or a static force applied at an eccentricity] at each story for each eccentricity to approximate these effects. Then static + dynamic response is combined. This is where we need to amplify accidental torsion. Because accidental torsion has not been calculated dynamically.
3. An "efficient and practical" approach is adopted in ETABS. After the analysis of MRSA cases;
a. Acceleration at each node is multiplied by tributary mass and given eccentricity so the result is a torsional force = m.a.e = F.e
b. A static response is generated under these torsional loads and added to MRSA dynamic results.
Now, strictly speaking, eccentricity was not directly analyzed in dynamic analysis in method 3 but atleast there is a satisfying globally used practice. So we dont need to increase accidental torsion by this method.
 
4. There is another method developed by  Fahjan et al. and quoted by CSi wiki. It also contains good background information. You can view the paper from csi wiki page.

Depth will help for the case of moment frame. Width can have a very minor to negligible effect as contribution of beam width to moment of inertia is small.

Hi Arslan, just browsing through your posts. You might already have the answer to this. To check if the columns are able to resist atleast 25% of the lateral force, you should do the following:
- Go to Mass Source and change the factors to 0.25 instead of 1. This would reduce the seismic base shear to 25%.
- Change your shear wall f12 modifier to nearly 0, i.e. 0.0001. This would mean that your shear walls will not resist any lateral forces. That is, all the lateral force will be resisted by columns only.
- Check if your columns are able to resist the lateral loads. If reinforcement in your columns is not exceeding 4%, you're good. 
- I would save this as a separate model and provide the maximum reinforcement in columns from this model and the original one. 

This topic troubled me a lot the last few months so I came across this post while doing my research. A quick update to the consideration of accidental torsion effects in RSA: the dynamic mass shifting procedure has been prohibited by Supplement # 2 to ASCE 7-16. Latest research found that buildings designed with this procedure were more susceptible to collapse.
Now we're only left with the static method of accounting for accidental torsional effects in RSA (including amplification). To accomplish this in ETabs, we can find the amplified eccentricities along both X and Y axes from the regular ELF analysis. These amplified eccentricities are input in the RS-X and RS-Y load cases respectively. I confirmed this approach with Dr. Justin Marshall, the co-author of "Guide to Seismic Provisions of ASCE 7-16", and Engr. Aung, Director AIT Solutions. 
As WR-1 said, Etabs (link) will obtain accidental torsional moments from story forces obtained from the combined RSA, then add these forces to the combined RSA results. This is the procedure prescribed in the "Guide to Seismic Provisions of ASCE 7-16" as well, snippet attached.
Omission of consideration of amplified accidental torsional effects in RSA results in very weak designs, and this is prescribed in both ASCE and UBC-97.
Note: the dynamic mass shifting analysis is still applicable for linear response history analysis. However, for response spectrum analysis it is prohibited.
 

Yes, it is. You can also release rotational restraint at the the end of landing beam.
 

Depth will help for the case of moment frame. Width can have a very minor to negligible effect as contribution of beam width to moment of inertia is small.

Wajahat,
I don't use American Codes anymore. You can check them to see what the requirements are. P-Delta analysis in my opinion, should be included irrespective of what code says.
This is code specific. For UBC, yes, for other please check the relevant code. If P-Delta is increasing your drifts to the extent that avoiding it is making your results pass then I guess your structure is super slender and you should beef up your framing.
I am note sure which 1.5 are you referring to. For stiffness modifiers, there are some recent discussions on the forum that you can find helpful. I am a bit rusty on them and apparently new ACI has changed a few things with regards to stiffness modifiers.
Thanks.

Unfortunately in Middle east i have experienced something similar, the person taking interview might not necessarily know all these answers as well but he will come up with these questions just to make sure that you know nothing and he knows everything, a bit unprofessional attitude, the questions should be as per the experience of the guy and the job description. I found out the reason for the interviews to be like that they prefer the reference alot rather going through interview process. Its just what i have seen definitely not same for all organisations.  

That is a well put response. On which forum did you get this response from?

Yes, the thickness of wall is appropriate as per your results; you need to provide the prescribed confinement at the ends of wall where compressive stresses are exceeding the limit set by the code.

You should make the judgment about the degree of rotational restraint available at the ends which can be based on calculations, intuition or experience. It depends on the end conditions:
single span with both ends of beam supported on masonry >>> no rotational restraint beam's end is supported on RCC columns>>> rotational restraint is available and depends on the relative stiffness of column to beam; one can use software, moment distribution method (MDM), or experience to get the distribution of moment between support and midspan. beam's end is continuous over support, meaning there is another beam on adjacent span>>> rotational restraint is available and depends on the span and depth of adjacent beam; again use software, MDM or moment coefficients from ACI or other similar source for moment coefficients.  

Looking at the results of mid-span and at interior support, a difference of 2% in results is not "significant" at all. You have not done anything wrong.

Here is a response that I received from Engr. Aung, Director AIT Solutions. Sharing for everyone's knowledge.
"You can consider the podium, Tower 1 and Tower 2 to be three different "towers", using multiple towers option. T3 is podium (Foundation to 3rd floor), T1 and T2 are Tower 1 and 2, starting from 4th floor to roof.
  However, it is suggested NOT to use ELF base shear from analysis if you have multiple towers in the model.   Normally, ELF base shear for each tower above the podium is calculated manually for scaling purpose only. To determine the base shear for scaling of two towers resting on a common podium, it is suggested to compute the design base shear of each tower above the podium (using weight of tower above podium) based on code-specified equations manually. Then, scale the base shear of each tower above the podium in a 2-tower combined model in different response spectrum cases. You will have two sets of response spectrum cases: RSX-T1 and RSY-T1; RSX-T2 and RSY-T2. Use 2-tower combined model to design each tower with corresponding response spectrum cases. For podium design, you may use the envelope of those cases.   For application of wind load in multiple towers, it is suggested to turn on "Allow multiple towers" in the Option menu and assign the diaphragm names separately for each tower. If the story height of each story is different between two towers, it is suggested to crosscheck the wind base shear of each tower above the podium with manual calculated results.   It is okay to connect two towers on a common podium. It is suggested to check the in-plane forces in podium diaphragm due to movement of towers under lateral loads.   If the seismic gap is allowed at the amenity podium floors and no constructability and maintenance issues, you can also provide the seismic gap and analyze the towers separately."

Strong column, weak beam. @Badar (BAZ) is our expert on this topic. I believe he can shed some light on beam column ratio etc.
 

Thank you for updating on this. If someone else gets into the same issue in the future, your post would be of great help to them.

Please solve one frame manually with these conditions ,you will understand

@UmarMakhzumiThank you for your reply.
I ended up finding it in edit => edit shell => reverse wall local 3 axis

 
 

Hello,
I am working on the design of a building as per ASCE 7-10 code. This is the first time I am working with this code. The codes I usually use are the Eurocodes.
The building is located in a seismic region.
My question is, does the columns bending moment capacity need to be equal to or superior then 1.3 times the capacity of the beams they support ?
This is an important rule in the Eurocodes but I can't seem to find it in the ASCE 7-10.
Many thanks.
 

Hello,
What I found is that the issue comes up only when I work with Microsoft OneDrive activated. When I close Microsoft OneDrive and run the model, it works normally. This is probably the reason why etabs technical support wasn't able to reproduce the error.
Thank you all for your replies and help.

Salam, Muhammad, in interviews you can expect anything, some of these questions are too bookish but not that hard if you are working in a structural design office.

Assalam O Alykum 
I gave very hard interview questions (at least for me )in famous international engineering consultancy on Microsoft team. Might be it looks very easy for many of you. 
Just want to share my experience.  If seniors can have a look on the questions which i remember and my replies to them. I believe should not ask with a guy having less experience. However, I am expecting to receive positive response.
Questions are underlined ,but not in sequence. 
Thank you  
Interview q.pdf

*SEFP Consistent Design*
*UBC Seismic Drift Limits*
*Doc No: 10-00-CD-0003*
*Date: June 04, 2013*
 
The goal of this tutorial is to demonstrate how to evaluate building drifts and story drifts using UBC 97 guidelines. The philosophy behind Story Drift Limits is “Deflection Control”; In UBC 97, deflection control is specified in terms of the story drift as a limit on the lateral displacement of one level relative to the level below. The story drift is determined from the maximum inelastic response, ΔM.
 
Let’s start by defining the design-level response displacements. The elastic deflections due to strength-level design seismic forces are called design-level response displacements. These are denoted by ΔS, where the subscript ‘s’ stands for strength design. Design level response displacements are what you get out of your software, when you run analysis. Please note that structural analysis softwares may provide these values in different formats; say a percentage of height or a direct output.
 
Well, to calculate your story drifts, first you need to find maximum inelastic response displacements from your design-level response displacements. The maximum inelastic response displacement is defined as:
ΔM = 0.7RΔS
Where, R is the structural system coefficient, the subscript ‘m’ in ΔM signifies that we are calculating a maximum value for the deflection due to seismic response that includes inelastic behavior.
 
Seismic drift values are much larger than wind values. UBC uses maximum inelastic response displacements rather than the design level displacements to verify the performance of the building. Seismic drift limits are 2% & 2.5% of the story height for long and short -period buildings. For a floor to floor height of 12 feet the max., allowable inelastic drift value would be 2% of 12 feet= 0.02*12*12 in=2.88 in. For wind for a 12 story height, drift would be L/400=12*12/400 =0.36 inches, A comparison of both wind and seismic drift limits shows that earthquake inelastic displacements are quiet large compared to wind displacements. That is why proper detailing is emphasized in seismic design.
 
When calculating ΔS for seismic, make sure:
You have included accidental torsion in your analysis. Use strength design load combinations: 1.2D + 1.0E + 0.5L & 0.9D + 1.0E. You are using cracked section properties for reinforced concrete buildings. Typical values are Icr walls= 0.5EcIg, Beams = 0.5EcI g & for Columns 0.5 - 0.7 EcIg. Use a reliability/ redundancy factor= 1 to calculate seismic forces. Whenever the dynamic analysis procedure of §1631 is used, story drift should be determined as the modal combination of the story drift for each mode. Determination of story drift from the difference of the combined mode displacements may produce erroneous results because maximum displacement at a given level may not occur simultaneously with those of the level above or below. Differences in the combined mode displacements can be less than the combined mode story drift.  
Example:
A four-story special moment-resisting frame (SMRF) building has the following design level response displacements.(See attached Image)

R= 7.0,
I= 1
Time period= 0.6 sec
(See the attached image for Story Information)
 
 
Calculate:
Maximum Inelastic response displacements. Story drift in story 3 due to ΔM. Check story 3 for story drift limit. Maximum Inelastic response displacements ΔM = 0.7RΔS
ΔM = (0.7) (7) ΔS = (4.9) ΔS
(See the attached image for Maximum Inelastic response displacements)

 
Story drift in story 3 due to ΔM Story 3 is located between Levels 2 and 3. Thus
ΔM drift = 5.39 - 3.43 = 1.96 in.
 
Check story 3 for story drift limit. For structures with a fundamental period less than 0.7 seconds, §1630.10.2 requires that the ΔM story drift not exceed 0.025 times the story height. For story 3:
Story drift using ΔM = 1.96 in.
Story drift limit = 0.025 *(12*12) in = 3.6 in. > 1.96 in. Therefore, Okay.

Metal Deck profile P-3623 is Zamil Steel deck. It has an in built shear studs, so yes it is a composite decking, which can the be supported through non-composite steel joists. Hope this an attached file clear the issue
ZS-PEB-Steel-Deck.pdf