Asif Mostafa

*Comments/Observations regarding modelling in ETABS*
*Doc No: 10-00-CD-0006*
*Date: May 06, 2017*
Some of the observations made during extraction of results from ETABS (v 9.7.4), for design of reinforced concrete members, are being share in this article.,
1) Minimum Eccentricity
ETABS always considers the minimum eccentricity for selecting the design moment of columns irrespective of the probable behavior of the column, whether short or long column. See section 10.10.6.5 and its commentary of ACI 318-08 which deals with minimum eccentricity of long columns. You should always check the design moments that ETABS uses for columns if you want to bring down the cost of construction.
2) Unbraced/ Braced Preference
If your model has lateral loads, ETABS will give you design moments in column irrespective of its status as braced or un-braced as per ACI 318 criteria. You should investigate if the storey under consideration is braced, or un-braced (10.10.5.2), and decide appropriate design moments of columns.
3) Time Period
ETABS has a tendency to select a time period of the building that is considerably less than the value obtained by the approximate method, Method A, of the section 1630.2.2  of UBC 97. To quote the FEMA 451 document: ''Because this formula is based on lower bound regression analysis of measured building response in California, it will generally result in periods that are lower (hence, more conservative for use in predicting base shear) than those computed from a more rigorous mathematical model". So, there is no need to use the value of time period that is lot less than Ta. One should always check the time period used by the software; ETABS can overestimate the seismic force by more than 2 times.
Method A gives lower T and higher V, so FEMA 451 has advised not to use the value of time period less than this value even if rigorous analysis gives a lower value.
I have seen the results where Etabs have use the value of time period less than Ta; in-fact as low as 0.5Ta, which can increase the base shear two times. (For a complete discussion on time period, please see the following this thread that complements this section).
4) Stiffness Modifiers
First thing is related to modelling the bending stiffness of flexural members, for strength level loads, that is representative of their condition near failure. The ACI code specifies the modifier of 0.35 on gross moment of inertia to represent its condition at yielding. 
Some people say that the factor should be multiplied by 2 to represent the stiffness of T-beam. This approach would be justified if you are not taking into the account the out of plan bending stiffness of slab.
But, ETABS does include the out of plane bending stiffness if you have modelled the slab by using shell elements. So, a factor of 0.7 would overestimate the stiffness of your structure in this case, and will lead to under-design.
If one has used the modifier of 0.35 in ETABS for beams in beam-slab floor system, then what value should be adopted for slab? It should not be 0.25, as this value has been specified for flat plates and flat sab floor system.
If one is using some value of modifier for out of plane bending stiffness on shells, then the share of the bending moment in beams will be reduced accordingly. This approach is correct if one will be providing the reinforcement in column strips of slab. But, if you are providing reinforcement in slab in the direction perpendicular to supports only, i.e. beams, as is the general practice in Pakistan, then you are under-estimating the flexural demand in beams.
Now, there is also a question of factors to be used while deciding the amount of reinforcement required in beams, columns and shear walls.
If you are using factors 0.35 for beams and shear walls, and 0.7 for columns, then you are finding out the demand in members at the point of yielding, and this conforms to the code. But, this also means that the structure might experience unacceptable cracks widths. So, if you are using 0.35 for calculating the demand at strength-level forces, then you should also perform crack-control-check at service-level loads by using the factor of 1.
If you are calculating the strength-level demand with a modifier of 1 for all structural members, after you have decided the location and the number of shear walls with modifier of 0.35, then you are overestimating seismic forces, as you are underestimating the time-period. But, the structural performance will improve.
This article is based on my two separate posts regarding the subject matter. You can view the discussion on the items raised above by viewing the following links:
1) http://www.sepakistan.com/topic/2008-issues-in-etabs-results/
2) http://www.sepakistan.com/topic/2290-modelling-issuesconsideration-in-etabs/
Thanks.

Thanks for your response and insight. But the plastic hinge is formed after a plastic deformation occurs. And at that point the beam do not abide by the Hook's Law. Hence you have to calculate and define the non linear properties for the materials and section along with linear properties.
By the way, is your goal is to performance evaluation due to seismic excitation? 

Thanks for your help.

I am working with Austrailian/New Zealand codes. (AS/NZS 1170). the vibrations come under the serviceablilty requirements. If you see the table attached we use 3kN superimposed load for the design check which reduced by 0.5times for the service check for the beam vibrations. AS/NZS 1170 says that minimum 1kN load should be checked for vibration with 1 to 2mm deflection under the point load but we use 1.5kN minimum. have a read on the attached as they are self explainatory.
 

Hi dear engineers, it's been a while I joined the forum. Currently, I am working as an Asst Senior Structural Engineer in ADCL, an affiliated partner of ADGI, USA. I am preparing myself for higher study in conjunction with my job. I have an immense interest in Performance-Based Seismic Design and Steel Structures. I always try to find literature, software, seminar notes, etc regarding these. Looking forward to enhancing my knowledge through mutual sharing.

Hi Asif,
Glad to have you on the forum. Welcome aboard.
Thanks.

Hi dear engineers, it's been a while I joined the forum. Currently, I am working as an Asst Senior Structural Engineer in ADCL, an affiliated partner of ADGI, USA. I am preparing myself for higher study in conjunction with my job. I have an immense interest in Performance-Based Seismic Design and Steel Structures. I always try to find literature, software, seminar notes, etc regarding these. Looking forward to enhancing my knowledge through mutual sharing.

FEMA P-2012/2018 Assessing Seismic Performance of Buildings with Configuration Irregularities is now available.
Thanks.
FEMA_P-2012_508.pdf

Source: https://www.fema.gov/media-library-data/1551300980344-8e2d825576db50c85ea48448ede5bd90/FEMA_P-2012_508.pdf
 

There is a rule of thumb that the Structural Engineers of 70's used and still used as a means of getting an idea about how steel may require. It is expressed as (story/3+7) psf.