Badar (BAZ)

From the seismic point of view, the former is recommended by the code (ACI 318-19). See the attached:

Thanks @Badar (BAZ)

From the seismic point of view, the former is recommended by the code (ACI 318-19). See the attached:

You should detail the side as per the maximum demand on either side.

You should make the judgment about the degree of rotational restraint available at the ends which can be based on calculations, intuition or experience. It depends on the end conditions:
single span with both ends of beam supported on masonry >>> no rotational restraint beam's end is supported on RCC columns>>> rotational restraint is available and depends on the relative stiffness of column to beam; one can use software, moment distribution method (MDM), or experience to get the distribution of moment between support and midspan. beam's end is continuous over support, meaning there is another beam on adjacent span>>> rotational restraint is available and depends on the span and depth of adjacent beam; again use software, MDM or moment coefficients from ACI or other similar source for moment coefficients.  

Salam, No engineering problem has single right solution, there are multiple equally good solutions for engineering problems. So difference of opinion on this matter is totally justified. Engineers decide as per their experience and engineering judgement supported by engineering theories.
I second Badar.
Regards

You should make the judgment about the degree of rotational restraint available at the ends which can be based on calculations, intuition or experience. It depends on the end conditions:
single span with both ends of beam supported on masonry >>> no rotational restraint beam's end is supported on RCC columns>>> rotational restraint is available and depends on the relative stiffness of column to beam; one can use software, moment distribution method (MDM), or experience to get the distribution of moment between support and midspan. beam's end is continuous over support, meaning there is another beam on adjacent span>>> rotational restraint is available and depends on the span and depth of adjacent beam; again use software, MDM or moment coefficients from ACI or other similar source for moment coefficients.  

Excellent brother
Procedure attached as per CSI
Procedure long term deflection.pdf

I think you are not getting his point. He is saying that if you allow column strip deflection to be 39mm as per L/240 criteria of the code, then that deflection will be translated to a value of 52mm at the mid span of middle strip.
Explanation: For 39mm deflection, the slope of the deflected slab will be 9.3/2/.039 = 119.5 (about 1 in 120); and based on that slope, the midspan deflection will be about 12.46/2/.120 = 51.9mm.
If possible, can you share how are you calculating your long term deflections of two way slab panels?

I think you are not getting his point. He is saying that if you allow column strip deflection to be 39mm as per L/240 criteria of the code, then that deflection will be translated to a value of 52mm at the mid span of middle strip.
Explanation: For 39mm deflection, the slope of the deflected slab will be 9.3/2/.039 = 119.5 (about 1 in 120); and based on that slope, the midspan deflection will be about 12.46/2/.120 = 51.9mm.
If possible, can you share how are you calculating your long term deflections of two way slab panels?

Read it again, you will change your your mind.  Read the note 1 of the ACI's table 8.3.1.1 (318-14) meant for the guidance regarding the minimum thickness  for non-prestressed slabs.

For the panel 2, I would go to table 7.9.2c of ACI 314R-11. The size of panel 2 will have to be determined from the outer most columns. Panel 1 is a "cornel panel".

The purpose of structural analysis, as Graham Powel puts it esoterically and with simplicity, "is not to get an exact simulation of behavior ..... the goal is to get demand/capacity ratios that are accurate enough for decision making". The exact analysis is not required for engineering problems, and it is not possible as well.
For most RC slab-beam systems, the slab is only required to resist gravity loads. For gravity loads, you do not need to do ELF. The coefficients method of ACI 314R-11 will do the job even with its limitations. Same is true for one-way panels; you can use coefficients in ACI 318.
For academic purposes, you can model your floor system in SAFE or can get the results for ETABS as well, as it can do design of slab as well.
In actual practice, the level of investigation and scrutiny put on a particular structural member depends on the importance of the structural member in overall structural system.
In your case, RC floor is the structural member with most load paths. It can transfer loads in ways unimaginable by most structural engineers. It does not warrant a detailed scrutiny unless it is acting as s transfer member at podium levels of a tall building.

Read it again, you will change your your mind.  Read the note 1 of the ACI's table 8.3.1.1 (318-14) meant for the guidance regarding the minimum thickness  for non-prestressed slabs.

Combination of following: intuition, observation, reading and talking about it.

Salam,
Adding my cent to the BAZ's response, intuition: your basic understanding, visualization, imagination, observation: observe different structural system around you, man-made and natural, observe how they sustain different kind of loads successfully, observe how seniors are tackling different problems, observe and investigate failures of structural members or structures, imagine how structure will fail  reading: read about structural materials, understand their properties and behaviors, weaknesses and benefits, understand their mechanics under loads, read structural analysis and structural design books talking: talking or thinking aloud, discussing with colleagues and fellow structural engineers, this builds confidence and clarity in concepts, if you can not talk about a concept its understanding is unclear yet.
 
@Badar (BAZ) please correct if anything added wrong.
 
Regards
 
 
 

I agree with this statement with a slight modification: "until unless you work with some competent engineers and challenge yourself with technical problems on bigger projects" which challenge your pre-existing pool of knowledge. 
But that statement covers the development of competency of individual. In order to land a job which pays well, this is not enough at all.
The MS degree has a huge advantage. A MS degree from a university of good standing in US, UK, Canada, AUS, New Zealand makes it easier for you to acquire a job at a company of international repute, and you are paid much better than the others as well.
In Gulf, one will routinely come across adds which requires that the applicant must be registered as professional structural engineer (like CEng, MIStructE) in an internationally recognized regulatory authority (mostly UK). These are the people which pay almost twice to the people that satisfy their requirement. The requirement is simple, they pay to the privileged ones.  This criteria of hiring anyone who is registered with a particular organization ( mostly Institute of Structural Engineers, UK) gives undue advantage; but, it is convenient for recruiters. In most cases, the one who is registered with Institute of Structural Engineers, would have got his Masters from there as well.
Any one who has got the job (especially in Structural engineering) in western firms based in Gulf without the MS from above mentioned countries, count your self extremely lucky.

Location of Base for Seismic Design.pdf
Read the attached document.
You initial design should not alter too much by incorporating the concern of third party in your model.

Location of Base for Seismic Design.pdf
Read the attached document.
You initial design should not alter too much by incorporating the concern of third party in your model.

*Comments/Observations regarding modelling in ETABS*
*Doc No: 10-00-CD-0006*
*Date: May 06, 2017*
Some of the observations made during extraction of results from ETABS (v 9.7.4), for design of reinforced concrete members, are being share in this article.,
1) Minimum Eccentricity
ETABS always considers the minimum eccentricity for selecting the design moment of columns irrespective of the probable behavior of the column, whether short or long column. See section 10.10.6.5 and its commentary of ACI 318-08 which deals with minimum eccentricity of long columns. You should always check the design moments that ETABS uses for columns if you want to bring down the cost of construction.
2) Unbraced/ Braced Preference
If your model has lateral loads, ETABS will give you design moments in column irrespective of its status as braced or un-braced as per ACI 318 criteria. You should investigate if the storey under consideration is braced, or un-braced (10.10.5.2), and decide appropriate design moments of columns.
3) Time Period
ETABS has a tendency to select a time period of the building that is considerably less than the value obtained by the approximate method, Method A, of the section 1630.2.2  of UBC 97. To quote the FEMA 451 document: ''Because this formula is based on lower bound regression analysis of measured building response in California, it will generally result in periods that are lower (hence, more conservative for use in predicting base shear) than those computed from a more rigorous mathematical model". So, there is no need to use the value of time period that is lot less than Ta. One should always check the time period used by the software; ETABS can overestimate the seismic force by more than 2 times.
Method A gives lower T and higher V, so FEMA 451 has advised not to use the value of time period less than this value even if rigorous analysis gives a lower value.
I have seen the results where Etabs have use the value of time period less than Ta; in-fact as low as 0.5Ta, which can increase the base shear two times. (For a complete discussion on time period, please see the following this thread that complements this section).
4) Stiffness Modifiers
First thing is related to modelling the bending stiffness of flexural members, for strength level loads, that is representative of their condition near failure. The ACI code specifies the modifier of 0.35 on gross moment of inertia to represent its condition at yielding. 
Some people say that the factor should be multiplied by 2 to represent the stiffness of T-beam. This approach would be justified if you are not taking into the account the out of plan bending stiffness of slab.
But, ETABS does include the out of plane bending stiffness if you have modelled the slab by using shell elements. So, a factor of 0.7 would overestimate the stiffness of your structure in this case, and will lead to under-design.
If one has used the modifier of 0.35 in ETABS for beams in beam-slab floor system, then what value should be adopted for slab? It should not be 0.25, as this value has been specified for flat plates and flat sab floor system.
If one is using some value of modifier for out of plane bending stiffness on shells, then the share of the bending moment in beams will be reduced accordingly. This approach is correct if one will be providing the reinforcement in column strips of slab. But, if you are providing reinforcement in slab in the direction perpendicular to supports only, i.e. beams, as is the general practice in Pakistan, then you are under-estimating the flexural demand in beams.
Now, there is also a question of factors to be used while deciding the amount of reinforcement required in beams, columns and shear walls.
If you are using factors 0.35 for beams and shear walls, and 0.7 for columns, then you are finding out the demand in members at the point of yielding, and this conforms to the code. But, this also means that the structure might experience unacceptable cracks widths. So, if you are using 0.35 for calculating the demand at strength-level forces, then you should also perform crack-control-check at service-level loads by using the factor of 1.
If you are calculating the strength-level demand with a modifier of 1 for all structural members, after you have decided the location and the number of shear walls with modifier of 0.35, then you are overestimating seismic forces, as you are underestimating the time-period. But, the structural performance will improve.
This article is based on my two separate posts regarding the subject matter. You can view the discussion on the items raised above by viewing the following links:
1) http://www.sepakistan.com/topic/2008-issues-in-etabs-results/
2) http://www.sepakistan.com/topic/2290-modelling-issuesconsideration-in-etabs/
Thanks.

I have done a lot of these using Canadian Code.
In older versions of ACI, Appendix D in ACI 318 used to cover all the limit states (mentioned by Baz above as well). I don't know what is in the latest version but you calculate your loads including moments if any, and then calculate the minimum embedment capacity for different limit states against applied loading using ACI Appendix D.
If you are using Hilit, they have a free software called Profis Anchor that can do all the calculations for you. Just download and provide your rebar configurations, concrete edge distances and loads and it will calculate the embedment using Hilti Products.
Thanks.

You need to follow the guidelines of chapter 17 of AC318-14/19 or equivalent sections from older versions of the same code. Assuming, you do not have tension in the column, your post installed rebars must have enough embedment to satisfy the limit states corresponding to steel failure, concrete pry out and concrete breakout failure modes.

There are typical details of reinforcement, which can be found in various detailing manuals, that the design engineers follow around the world. They do not normally check development length for typical/usual member sizes with usual/ conventional loadings, instead they follow those reinforcement details.
 
The critical section is the section with maximum tensile stress. For most beams or flexural members, it is the face of column or support.
 
For this particular case, you can treat the wall-end as simply supported. But, there are other ways to develop the reinforcement such as mechanical anchorages (headed bar) or welding with end-plates.
 
As you can see in your attached diagram, the length of hook is not considered by American codes. Some codes do consider them explicitly. For American codes, any length beyond 12db does not contribute.

By joint translation, you mean that the manual calculations were based on the assumption that frame can not move side ways?
Anyways, for practical purposes, this is not a significant difference.

You do not need any complicated detail for that. Just leave the top bars without hook in beam-column joint.